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HomeMy WebLinkAboutDOCUMENT Engineering Calcs stamped 1701 North Lake Ave 2020-03-30WD\CALC\CALCMSTR ASTRA Engineering 7887 E. Belleview Avenue, Suite 1100 Englewood, CO 80111 JOB NO. 18-170 DATE 3-27-20 revised DESIGNED BY: CHECKED BY: PAGE CLIENT: ESA Associates, P.C. 1919 Seventh St. Boulder, CO 80302 PROJECT: Rocky Mountain Hotel Remodel (Supplemental) 1701 North Lake Ave. Estes Park, CO DESIGN CRITERIA AND MATERIALS DESIGN DATA CODE DATA CONCRETE: Type of Concrete F’c (psi) BUILDING CODE: 2015 IBC Slabs on Grade 4000 w/ City/County Amendments Footings Grade Beams 4000 All Others 3000 SEISMIC CRITERIA: Design Category: “B” STEEL ASTM A 615, Grade 60 for all Reinforcing Site Classification: “D” REINFORCING: ASTM A 706, Grade 60 for all Welded Bars Ss: 0.211g ASTM A185, Wire per ASTM A82 all Welded Wire Fabric S1: 0.063g Use Group: 1 SOIL: Soil Report By: IBC Table 1804.6 Soil Report No: Bearing Pressure: 3000 PSF WIND CRITERIA: Wind Speed Vall. = 124 mph STRUCTURAL Wide Flange Members - ASTM A992 (Fy = 50 ksi) I = 1.0 STEEL: Channels, Plates & Angles – ASTM A36 (Fy = 36 ksi) Exposure = “C” Pipe Steel – ASTM A53, Gr.-B (Fy = 35 ksi) Tube Steel – ASTM A500, Gr-B (Fy = 46 ksi) DESIGN DEAD LOADS: Wood Deck – 2.0 psf Trusses – 3.5 psf MASONRY: ASTM C90, Gr. N, Type I, F’m = 1900 psi MEP – 6.5 psf Insulation –4.5 psf Finish Materials – 2.0 MORTAR: Type S, 1800 psi Misc. – 1.5 TOTAL ROOF (DL): 20 PSF -------------------------------------- MASONRY 2000 psi Concrete – 44.0 psf GROUT: MEP – 6.5 psf Insulation – 4.5 psf Joist – 3.5 psf WOOD: Wood Type Fb (psi) Fv (psi) E (ksi) Fc (psi) Finish Materials – 2.0 psf Misc. – 14.5 psf Joist, Ledgers, Plates 4x or Smaller (UNO) 1000 150 1600 1300 TOTAL FLOOR (DL): 75 PSF DESIGN LIVE LOADS: Flat Roof Min. Snow – 70 psf Floor Live Load – 40 PSF Floor Live Load Public Area-100 PSF Canopy Snow Load – 70 PSF Stair & Corridor Live Load – 100 PSF Project Job Ref. Section Wind Force (revised) Sheet no./rev. 1 Calc. by N Date 3/26/2020 Chk'd by Date App'd by Date WIND LOADING In accordance with ASCE7-10 Using the directional design method Tedds calculation version 2.1.03 50 ft 3 2 f t Plan 32 ft 28 f t Elevation Building data Type of roof;Arched Length of building;b = 50.00 ft Width of building;d = 32.00 ft Height to eaves;H = 24.00 ft Radius of arch;rarc = 34.0 ft Diameter of arch;Darc = 68.0 ft Rise of arch;Harc = 4.00 ft Angle of roof at springer;eaves = 28.1 deg Mean height;h = 26.00 ft General wind load requirements Basic wind speed;V = 160.0 mph Risk category;II Velocity pressure exponent coef (Table 26.6-1);Kd = 0.85 Exposure category (cl 26.7.3);C Enclosure classification (cl.26.10);Partially enclosed buildings Internal pressure coef +ve (Table 26.11-1);GCpi_p = 0.55 Internal pressure coef –ve (Table 26.11-1);GCpi_n = -0.55 Gust effect factor;Gf = 0.85 Minimum design wind loading (cl.27.4.7);pmin_r = 8 lb/ft2 Topography Topography factor not significant;Kzt = 1.0 Velocity pressure equation;q = 0.00256  Kz  Kzt  Kd  V2  1psf/mph2; Velocity pressures table z (ft) Kz (Table 27.3-1) qz (psf) 15.00 0.85 47.35 15.00 0.85 47.35 Project Job Ref. Section Wind Force (revised) Sheet no./rev. 2 Calc. by N Date 3/26/2020 Chk'd by Date App'd by Date z (ft) Kz (Table 27.3-1) qz (psf) 20.00 0.90 50.14 24.00 0.93 51.92 26.00 0.95 52.81 28.00 0.96 53.70 Peak velocity pressure for internal pressure Peak velocity pressure – internal (as roof press.);qi = 52.81 psf Pressures and forces Net pressure;p = q  Gf  Cpe - qi  GCpi; Net force;Fw = p  Aref; Roof load case 1 - Wind 0, GCpi 0.55, -cpe Zone Ref. height (ft) Ext pressure coefficient cpe Peak velocity pressure qp (psf) Net pressure p (psf) Area Aref (ft2) Net force Fw (kips) A (-ve) 26.00 -0.90 52.81 -69.44 429.14 -29.80 B (-ve) 26.00 -0.82 52.81 -66.08 807.57 -53.36 C (-ve) 26.00 -0.50 52.81 -51.49 429.14 -22.10 Total vertical net force;Fw,v = -101.92 kips Total horizontal net force;Fw,h = -2.72 kips Walls load case 1 - Wind 0, GCpi 0.55, -cpe Zone Ref. height (ft) Ext pressure coefficient cpe Peak velocity pressure qp (psf) Net pressure p (psf) Area Aref (ft2) Net force Fw (kips) A1 15.00 0.80 47.35 3.15 750.00 2.36 A2 15.00 0.80 47.35 3.15 0.00 0.00 A3 24.00 0.80 51.92 6.26 450.00 2.82 B 26.00 -0.50 52.81 -51.49 1200.00 -61.79 C 26.00 -0.70 52.81 -60.47 854.39 -51.66 D 26.00 -0.70 52.81 -60.47 854.39 -51.66 Overall loading Projected vertical plan area of wall;Avert_w_0 = b  H = 1200.00 ft2 Projected vertical area of roof;Avert_r_0 = b  Harc = 200.00 ft2 Minimum overall horizontal loading;Fw,total_min = pmin_w  Avert_w_0 + pmin_r  Avert_r_0 = 20.80 kips Leeward net force;Fl = Fw,wB = -61.8 kips Windward net force;Fw = Fw,wA_1 + Fw,wA_2 + Fw,wA_3 = 5.2 kips Overall horizontal loading;Fw,total = max(Fw - Fl + Fw,h, Fw,total_min) = 64.3 kips Roof load case 2 - Wind 0, GCpi -0.55, -1cpe Project Job Ref. Section Wind Force (revised) Sheet no./rev. 3 Calc. by N Date 3/26/2020 Chk'd by Date App'd by Date Zone Ref. height (ft) Ext pressure coefficient cpe Peak velocity pressure qp (psf) Net pressure p (psf) Area Aref (ft2) Net force Fw (kips) A (-ve) 26.00 -0.90 52.81 -11.35 429.14 -4.87 B (-ve) 26.00 -0.82 52.81 -7.99 807.57 -6.45 C (-ve) 26.00 -0.50 52.81 6.60 429.14 2.83 Total vertical net force;Fw,v = -8.36 kips Total horizontal net force;Fw,h = -2.72 kips Walls load case 2 - Wind 0, GCpi -0.55, -1cpe Zone Ref. height (ft) Ext pressure coefficient cpe Peak velocity pressure qp (psf) Net pressure p (psf) Area Aref (ft2) Net force Fw (kips) A1 15.00 0.80 47.35 61.24 750.00 45.93 A2 15.00 0.80 47.35 61.24 0.00 0.00 A3 24.00 0.80 51.92 64.35 450.00 28.96 B 26.00 -0.50 52.81 6.60 1200.00 7.92 C 26.00 -0.70 52.81 -2.38 854.39 -2.03 D 26.00 -0.70 52.81 -2.38 854.39 -2.03 Overall loading Projected vertical plan area of wall;Avert_w_0 = b  H = 1200.00 ft2 Projected vertical area of roof;Avert_r_0 = b  Harc = 200.00 ft2 Minimum overall horizontal loading;Fw,total_min = pmin_w  Avert_w_0 + pmin_r  Avert_r_0 = 20.80 kips Leeward net force;Fl = Fw,wB = 7.9 kips Windward net force;Fw = Fw,wA_1 + Fw,wA_2 + Fw,wA_3 = 74.9 kips Overall horizontal loading;Fw,total = max(Fw - Fl + Fw,h, Fw,total_min) = 64.3 kips Roof load case 3 - Wind 90, GCpi 0.55, -cpe Zone Ref. height (ft) Ext pressure coefficient cpe Peak velocity pressure qp (psf) Net pressure p (psf) Area Aref (ft2) Net force Fw (kips) A (-ve) 26.00 -0.91 52.81 -69.88 433.12 -30.27 B (-ve) 26.00 -0.89 52.81 -69.08 433.12 -29.92 C (-ve) 26.00 -0.51 52.81 -51.85 799.61 -41.46 Total vertical net force;Fw,v = -97.63 kips Total horizontal net force;Fw,h = 0.00 kips Walls load case 3 - Wind 90, GCpi 0.55, -cpe Zone Ref. height (ft) Ext pressure coefficient cpe Peak velocity pressure qp (psf) Net pressure p (psf) Area Aref (ft2) Net force Fw (kips) A1 15.00 0.80 47.35 3.15 480.00 1.51 A2 20.00 0.80 50.14 5.05 160.00 0.81 Project Job Ref. Section Wind Force (revised) Sheet no./rev. 4 Calc. by N Date 3/26/2020 Chk'd by Date App'd by Date Zone Ref. height (ft) Ext pressure coefficient cpe Peak velocity pressure qp (psf) Net pressure p (psf) Area Aref (ft2) Net force Fw (kips) A3 28.00 0.80 53.70 7.47 128.00 0.96 B 26.00 -0.39 52.81 -46.44 854.39 -39.68 C 26.00 -0.70 52.81 -60.47 1200.00 -72.56 D 26.00 -0.70 52.81 -60.47 1200.00 -72.56 Overall loading Projected vertical plan area of wall;Avert_w_90 = 854.39 ft2 Projected vertical area of roof;Avert_r_90 = 0.00 ft2 Minimum overall horizontal loading;Fw,total_min = pmin_w  Avert_w_90 + pmin_r  Avert_r_90 = 13.67 kips Leeward net force;Fl = Fw,wB = -39.7 kips Windward net force;Fw = Fw,wA_1 + Fw,wA_2 + Fw,wA_3 = 3.3 kips Overall horizontal loading;Fw,total = max(Fw - Fl + Fw,h, Fw,total_min) = 43.0 kips Roof load case 4 - Wind 90, GCpi -0.55, -cpe Zone Ref. height (ft) Ext pressure coefficient cpe Peak velocity pressure qp (psf) Net pressure p (psf) Area Aref (ft2) Net force Fw (kips) A (-ve) 26.00 -0.91 52.81 -11.79 433.12 -5.11 B (-ve) 26.00 -0.89 52.81 -10.99 433.12 -4.76 C (-ve) 26.00 -0.51 52.81 6.24 799.61 4.99 Total vertical net force;Fw,v = -4.68 kips Total horizontal net force;Fw,h = 0.00 kips Walls load case 4 - Wind 90, GCpi -0.55, -cpe Zone Ref. height (ft) Ext pressure coefficient cpe Peak velocity pressure qp (psf) Net pressure p (psf) Area Aref (ft2) Net force Fw (kips) A1 15.00 0.80 47.35 61.24 480.00 29.40 A2 20.00 0.80 50.14 63.14 160.00 10.10 A3 28.00 0.80 53.70 65.56 128.00 8.39 B 26.00 -0.39 52.81 11.65 854.39 9.95 C 26.00 -0.70 52.81 -2.38 1200.00 -2.85 D 26.00 -0.70 52.81 -2.38 1200.00 -2.85 Overall loading Projected vertical plan area of wall;Avert_w_90 = 854.39 ft2 Projected vertical area of roof;Avert_r_90 = 0.00 ft2 Minimum overall horizontal loading;Fw,total_min = pmin_w  Avert_w_90 + pmin_r  Avert_r_90 = 13.67 kips Leeward net force;Fl = Fw,wB = 10.0 kips Windward net force;Fw = Fw,wA_1 + Fw,wA_2 + Fw,wA_3 = 47.9 kips Overall horizontal loading;Fw,total = max(Fw - Fl + Fw,h, Fw,total_min) = 37.9 kips Project Job Ref. Section Wind Force (revised) Sheet no./rev. 5 Calc. by N Date 3/26/2020 Chk'd by Date App'd by Date A B C8 f t 16 f t 8 f t 50 ft 32 f t Wi n d - 0 o Plan view - Arched roof C 32 ft 24 f t Side face A1 A2 A3 50 ft 24 f t Windward face B 50 ft 24 f t Leeward face Project Job Ref. Section Wind Force (revised) Sheet no./rev. 6 Calc. by N Date 3/26/2020 Chk'd by Date App'd by Date A B C 13 ft 13 ft 24 ft 50 ft 32 f t Wind - 90o Plan view - Arched roof C 50 ft 24 f t Side face A1 A2 A3 32 ft 24 f t Windward face B 32 ft 24 f t Leeward face Project Job Ref. Section Parapet Analysis Sheet no./rev. 1 Calc. by N Date 3/27/2020 Chk'd by Date App'd by Date STEEL MEMBER ANALYSIS & DESIGN (AISC 360) In accordance with AISC360 14th Edition published 2010 using the ASD method Tedds calculation version 4.3.04 ANALYSIS Tedds calculation version 1.0.27 Geometry Geometry (ft) - Steel (AISC) - C 3x6 Me m b e r 1 Me m b e r 2 12 25 1 2 3 X Z Materials Name Density Youngs Modulus Shear Modulus Thermal Coefficient (lbm/ft3) ksi ksi C-1 Steel (AISC) 490 29000 11200 0.000012 Sections Name Area Moment of inertia Shear area parallel to Major Minor Minor Major (in2) (in 4) (in 4) (in 2) (in 2) C 3x6 2 2 0 1 1 Nodes Node Co-ordinates Freedom Coordinate system Spring X Z X Z Rot. Name Angle X Z Rot. (ft) (ft)()(kips/ft) (kips/ft)kip_ft/ 1 0 0 Fixed Fixed Free 0 0 0 0 2 0 2 Fixed Fixed Free 0 0 0 0 Project Job Ref. Section Parapet Analysis Sheet no./rev. 2 Calc. by N Date 3/27/2020 Chk'd by Date App'd by Date Node Co-ordinates Freedom Coordinate system Spring X Z X Z Rot. Name Angle X Z Rot. (ft) (ft)()(kips/ft) (kips/ft)kip_ft/ 3 0 7 Free Free Free 0 0 0 0 Elements Element Length Nodes Section Material Releases Rotated (ft) Start End Start moment End moment Axial 1 2 1 2 C 3x6 Steel (AISC) Fixed Fixed Fixed 2 5 2 3 C 3x6 Steel (AISC) Fixed Fixed Fixed Members Name Elements Start End Member1 1 1 Member2 2 2 Loading Self weight included Wind - Loading (kips/ft) 0.12 0.12 0.12 0.12 Me m b e r 1 Me m b e r 2 X Z Load combination factors Load combination Se l f W e i g h t De a d Li v e Wi n d 1.0D + 1.0L (Strength) 1.00 1.00 1.00 Project Job Ref. Section Parapet Analysis Sheet no./rev. 3 Calc. by N Date 3/27/2020 Chk'd by Date App'd by Date Load combination Se l f W e i g h t De a d Li v e Wi n d 0.6D + 0.6W (Strength) 0.60 0.60 0.60 Member Loads Member Load case Load Type Orientation Description Member1 Wind UDL GlobalX 0.12 kips/ft Member2 Wind UDL GlobalX 0.12 kips/ft Results Forces Strength combinations - Moment envelope (kip_ft) -0.9-0.9 Project Job Ref. Section Parapet Analysis Sheet no./rev. 4 Calc. by N Date 3/27/2020 Chk'd by Date App'd by Date Strength combinations - Shear envelope (kips) -0.5 0.4 ; Safety factors Shear;v = 1.67 Flexure;b = 1.67 Tensile yielding;t,y = 1.67 Tensile rupture;t,r = 2.00 Compression;c = 1.67 Member1 design Section details Section type;C 3x6 (AISC 15th Edn (v15.0)) ASTM steel designation;User defined Steel yield stress;Fy = 33 ksi Steel tensile stress;Fu = 58 ksi Modulus of elasticity;E = 29000 ksi Project Job Ref. Section Parapet Analysis Sheet no./rev. 5 Calc. by N Date 3/27/2020 Chk'd by Date App'd by Date 1.6" 0.36" 3" 0. 2 7 " 0. 2 7 " C 3x6 (AISC 15th Edn (v15.0)) Section depth, d, 3 in Section breadth, bf, 1.6 in Weight of section, Weight, 6 lbf/ft Flange thickness, tf, 0.273 in Web thickness, tw, 0.356 in Area of section, A, 1.8 in2 Radius of gyration about x-axis, rx, 1.09 in Radius of gyration about y-axis, ry, 0.413 in Elastic section modulus about x-axis, Sx, 1.38 in3 Elastic section modulus about y-axis, Sy, 0.263 in3 Plastic section modulus about x-axis, Zx, 1.74 in3 Plastic section modulus about y-axis, Zy, 0.543 in3 Second moment of area about x-axis, I x, 2.07 in4 Second moment of area about y-axis, I y, 0.3 in4 Lateral restraint Both flanges have lateral restraint at supports only Consider Combination 2 - 0.6D + 0.6W (Strength) Classification of sections for local buckling - Section B4 Classification of flanges in flexure - Table B4.1b (case 10) Width to thickness ratio;bf / tf = 5.86 Limiting ratio for compact section;pff = 0.38  [E / Fy] = 11.26 Limiting ratio for non-compact section;rff = 1.0  [E / Fy] = 29.64;Compact Classification of web in flexure - Table B4.1b (case 15) Width to thickness ratio;(d - 2  k) / tw = 4.56 Limiting ratio for compact section;pwf = 3.76  [E / Fy] = 111.46 Limiting ratio for non-compact section;rwf = 5.70  [E / Fy] = 168.97;Compact Section is compact in flexure Classification of flanges in uniform compression - Table B4.1a (case 1) Width to thickness ratio;bf / tf = 5.86 Limiting ratio for non-compact section;rfc = 0.56  [E / Fy] = 16.60;Nonslender Classification of web in uniform compression - Table B4.1a (case 5) Width to thickness ratio;(d - 2  k) / tw = 4.56 Limiting ratio for non-compact section;rwc = 1.49  [E / Fy] = 44.17;Nonslender Section is nonslender in compression Check design at end of span Design of members for tension - Chapter D Required tensile strength;Pr = 0.004 kips Slenderness limitations - Section D1 Slenderness ratio; = max(Lb,x / rx, Lb,y / ry) = 145.278 Tension member slenderness ratio does not exceed recommended limit of 300 Project Job Ref. Section Parapet Analysis Sheet no./rev. 6 Calc. by N Date 3/27/2020 Chk'd by Date App'd by Date Allowable tensile strength - D2 Nominal tensile yielding strength;Pn,yld = Fy  A = 58.08 kips Nominal tensile rupture strength;Pn,r = Fu  Ae = 102.08 kips Allowable tensile strength;Pc = min(Pn,yld / t,y, Pn,r / t,r) = 34.778 kips Pr / Pc = 0 PASS - Nominal tensile strength exceeds required tensile strength Design of members for shear - Chapter G Required shear strength;Vr,x = 0.5 kips Web area;Aw = d  tw = 1.068 in2 Web plate buckling coefficient;kv = 5 (d - 2  k) / tw <= 1.10  (kv  E / Fy) Web shear coefficient - eq G2-3;Cv = 1.000 Nominal shear strength - eq G2-1;Vn,x = 0.6  Fy  Aw  Cv = 21.1 kips Safety factor;v = 1.67 Allowable shear strength;Vc,x = Vn,x / v = 12.7 kips Vr,x / Vc,x = 0.041 PASS - Allowable shear strength exceeds required shear strength Design of members for flexure - Chapter F Required flexural strength;Mr,x = 0.9 kips_ft Yielding - Section F2.1 Nominal flexural strength for yielding - eq F2-1;Mn,yld,x = Mp,x = Fy  Zx = 4.8 kips_ft Lateral-torsional buckling - Section F2.2 Unbraced length;Lb = 2 ft Limiting unbraced length for yielding - eq F2-5;Lp = 1.76  ry  (E / Fy) = 1.796 ft Distance between flange centroids ;ho = 2.73 in c = ho / 2  (Iy / Cw) = 1.1 rts = 0.519 in Limiting unbraced length for inelastic LTB - eq F2-6;Lr = 1.95  rts  E / (0.7  Fy)  ((J  c / (Sx  ho)) + ((J  c / (Sx  ho))2 + 6.76  (0.7  Fy / E)2)) = 21.811 ft Moment at quarter point of segment;MA = 0.2 kips_ft Moment at center-line of segment;MB = 0.4 kips_ft Moment at three quarter point of segment ;MC = 0.6 kips_ft Maximum moment in segment;Mmax = 0.9 kips_ft LTB modification factor - eq F1-1;Cb = 12.5  Mmax / (2.5  Mmax + 3  MA + 4  MB + 3  MC) = 1.746 Nominal flexural strength for lateral-torsional buckling - eq F2-2 Mn,ltb,x = min(Cb  (Mp,x - (Mp,x - 0.7  Fy  Sx)  (Lb - Lp) / (Lr - Lp)), Mp,x) = 4.8 kips_ft Allowable flexural strength - F1 Nominal flexural strength;Mn,x = min(Mn,yld,x, Mn,ltb,x) = 4.8 kips_ft Allowable flexural strength;Mc,x = Mn,x / b = 2.9 kips_ft Mr,x / Mc,x = 0.314 PASS - Allowable flexural strength exceeds required flexural strength Project Job Ref. Section Parapet Analysis Sheet no./rev. 7 Calc. by N Date 3/27/2020 Chk'd by Date App'd by Date Design of members for combined forces - Chapter H Combined flexure and axial force - eq H1-1b;Pr / (2  Pc) + Mr,x / Mc,x = 0.314 PASS - Combined flexure and axial force is within acceptable limits Check design 1ft 2.003in along span Design of members for x-x axis deflection Maximum deflection;x = 0.006 in Allowable deflection;x,Allowable = Lm1_s1 / 240 = 0.1 in x / x,Allowable = 0.062 PASS - Allowable deflection exceeds design deflection Member2 design Section details Section type;C 3x6 (AISC 15th Edn (v15.0)) ASTM steel designation;User defined Steel yield stress;Fy = 33 ksi Steel tensile stress;Fu = 58 ksi Modulus of elasticity;E = 29000 ksi 1.6" 0.36" 3" 0. 2 7 " 0. 2 7 " C 3x6 (AISC 15th Edn (v15.0)) Section depth, d, 3 in Section breadth, bf, 1.6 in Weight of section, Weight, 6 lbf/ft Flange thickness, tf, 0.273 in Web thickness, tw, 0.356 in Area of section, A, 1.8 in2 Radius of gyration about x-axis, rx, 1.09 in Radius of gyration about y-axis, ry, 0.413 in Elastic section modulus about x-axis, Sx, 1.38 in3 Elastic section modulus about y-axis, Sy, 0.263 in3 Plastic section modulus about x-axis, Zx, 1.74 in3 Plastic section modulus about y-axis, Zy, 0.543 in3 Second moment of area about x-axis, Ix, 2.07 in4 Second moment of area about y-axis, Iy, 0.3 in4 Lateral restraint Both flanges have lateral restraint at supports only Consider Combination 2 - 0.6D + 0.6W (Strength) Classification of sections for local buckling - Section B4 Classification of flanges in flexure - Table B4.1b (case 10) Width to thickness ratio;bf / tf = 5.86 Limiting ratio for compact section;pff = 0.38  [E / Fy] = 11.26 Limiting ratio for non-compact section;rff = 1.0  [E / Fy] = 29.64;Compact Classification of web in flexure - Table B4.1b (case 15) Width to thickness ratio;(d - 2  k) / tw = 4.56 Limiting ratio for compact section;pwf = 3.76  [E / Fy] = 111.46 USE 600S162-43 STEEL STUDS Sx =0.767 IN3 Ix = 2.316 IN4 Project Job Ref. Section Parapet Analysis Sheet no./rev. 8 Calc. by N Date 3/27/2020 Chk'd by Date App'd by Date Limiting ratio for non-compact section;rwf = 5.70  [E / Fy] = 168.97;Compact Section is compact in flexure Classification of flanges in uniform compression - Table B4.1a (case 1) Width to thickness ratio;bf / tf = 5.86 Limiting ratio for non-compact section;rfc = 0.56  [E / Fy] = 16.60;Nonslender Classification of web in uniform compression - Table B4.1a (case 5) Width to thickness ratio;(d - 2  k) / tw = 4.56 Limiting ratio for non-compact section;rwc = 1.49  [E / Fy] = 44.17;Nonslender Section is nonslender in compression Check design at start of span Design of members for compression - Chapter E Required compressive strength;Pr = 0 kips Slenderness limitations and effective length - Section E2 Unbraced length;Lb,x = Lm2_s1 = 5 ft Effective length factor;Kx = 1.00 Column slenderness;x = Kx  Lb,x / rx = 55.046 Major axis column slenderness ratio does not exceed recommended limit of 200 Slenderness limitations and effective length - Section E2 Unbraced length;Lb,y = Lm2_s1_seg1 = 5 ft Effective length factor;Ky = 1.00 Column slenderness;y = Ky  Lb,y / ry = 145.278 Minor axis column slenderness ratio does not exceed recommended limit of 200 Flexural buckling of members without slender elements - Section E3 Elastic critical buckling stress - eq E3-4;Fe,x = 2  E / (Kx  Lb,x / rx)2 = 94.5 ksi Flexural buckling stress - eq E3-2;Fcr,x = [0.658Fy / Fe,x]  Fy = 28.5 ksi Nominal compressive strength for flexural buckling - eq E3-1; Pn,fb,x = Fcr,x  A = 50.2 kips Elastic critical buckling stress - eq E3-4;Fe,y = 2  E / (Ky  Lb,y / ry)2 = 13.6 ksi Flexural buckling stress - eq E3-3;Fcr,y = 0.877  Fe,y = 11.9 ksi Nominal compressive strength for flexural buckling - eq E3-1; Pn,fb,y = Fcr,y  A = 20.9 kips Torsional and torsional-flexural buckling of members without slender elements - Section E4 Elastic torsional buckling stress - eq E4-9;Fe,z = (2  E  Cw / (Kz  Lb,z)2 + G  J) / (A  ro2) = 245.1 ksi Unbraced length;Lb,z = Lm2_s1_seg1_R = 5 ft Effective length factor;Kz = 1.00 Flexural-torsional elastic buckling stress - eq E4-5;Fe = (Fe,x + Fe,z) / (2  H)  [1 - (1 - 4  Fe,x  Fe,z  H / (Fe,x + Fe,z)2)] = 81.8 ksi Flexural-torsional buckling stress - eq E3-2;Fcr = [0.658Fy / Fe]  Fy = 27.9 ksi Nominal compressive strength for torsional and flexural-torsional buckling - eq E4-1 ; Pn,ftb = Fcr  A = 49.1 kips Project Job Ref. Section Parapet Analysis Sheet no./rev. 9 Calc. by N Date 3/27/2020 Chk'd by Date App'd by Date Allowable compressive strength - E1 Nominal compressive strength;Pn = min(Pn,fb,x, Pn,fb,y, Pn,ftb) = 20.9 kips Allowable compressive strength;Pc = Pn / c = 12.5 kips Pr / Pc = 0.001 PASS - Nominal compressive strength exceeds required compressive strength Design of members for shear - Chapter G Required shear strength;Vr,x = 0.4 kips Web area;Aw = d  tw = 1.068 in2 Web plate buckling coefficient;kv = 5 (d - 2  k) / tw <= 1.10  (kv  E / Fy) Web shear coefficient - eq G2-3;Cv = 1.000 Nominal shear strength - eq G2-1;Vn,x = 0.6  Fy  Aw  Cv = 21.1 kips Safety factor;v = 1.67 Allowable shear strength;Vc,x = Vn,x / v = 12.7 kips Vr,x / Vc,x = 0.028 PASS - Allowable shear strength exceeds required shear strength Design of members for flexure - Chapter F Required flexural strength;Mr,x = 0.9 kips_ft Yielding - Section F2.1 Nominal flexural strength for yielding - eq F2-1;Mn,yld,x = Mp,x = Fy  Zx = 4.8 kips_ft Lateral-torsional buckling - Section F2.2 Unbraced length;Lb = 5 ft Limiting unbraced length for yielding - eq F2-5;Lp = 1.76  ry  (E / Fy) = 1.796 ft Distance between flange centroids ;ho = 2.73 in c = ho / 2  (Iy / Cw) = 1.1 rts = 0.519 in Limiting unbraced length for inelastic LTB - eq F2-6;Lr = 1.95  rts  E / (0.7  Fy)  ((J  c / (Sx  ho)) + ((J  c / (Sx  ho))2 + 6.76  (0.7  Fy / E)2)) = 21.811 ft Moment at quarter point of segment;MA = 0.5 kips_ft Moment at center-line of segment;MB = 0.2 kips_ft Moment at three quarter point of segment ;MC = 0.1 kips_ft Maximum moment in segment;Mmax = 0.9 kips_ft LTB modification factor - eq F1-1;Cb = 12.5  Mmax / (2.5  Mmax + 3  MA + 4  MB + 3  MC) = 2.326 Nominal flexural strength for lateral-torsional buckling - eq F2-2 Mn,ltb,x = min(Cb  (Mp,x - (Mp,x - 0.7  Fy  Sx)  (Lb - Lp) / (Lr - Lp)), Mp,x) = 4.8 kips_ft Allowable flexural strength - F1 Nominal flexural strength;Mn,x = min(Mn,yld,x, Mn,ltb,x) = 4.8 kips_ft Allowable flexural strength;Mc,x = Mn,x / b = 2.9 kips_ft Mr,x / Mc,x = 0.314 PASS - Allowable flexural strength exceeds required flexural strength Project Job Ref. Section Parapet Analysis Sheet no./rev. 10 Calc. by N Date 3/27/2020 Chk'd by Date App'd by Date Design of members for combined forces - Chapter H Combined flexure and axial force - eq H1-1b;Pr / (2  Pc) + Mr,x / Mc,x = 0.315 PASS - Combined flexure and axial force is within acceptable limits Check design at end of span Design of members for x-x axis deflection Maximum deflection;x = 0.248 in Allowable deflection;x,Allowable = Lm2_s1 / 240 = 0.25 in x / x,Allowable = 0.992 PASS - Allowable deflection exceeds design deflection 33 Direct Fastening Technical Guide, Edition 18 * More details about the innovative X-P and X-U fasteners can be found in Section 3.2.6. Allowable loads in minimum f'c   Qsi structural lightweight concrete1,5 Fastener description Fastener Shank diameter in. (mm) Minimum embedment in. (mm) Fastener location Installed into concrete Installed through 3" deep metal deck into concrete2,3 Tension lb (kN) Shear lb (kN) Tension lb (kN)Shear lb (kN) Upper flute Lower flute Upper flute Lower flute Premium Concrete Fastener X-P*0.157 (4.0) 3/4 (19)155 (0.7)165 (0.7)130 (0.6)105 (0.5) 285 (1.3) 285 (1.3) 1 (25)225 (1.0)300 (1.3)215 (1.0)165 (0.7) 340 (1.5) 340 (1.5) 1-1/4 (32)325 (1.4)445 (2.0)295 (1.3)230 (1.0) 375 (1.7) 375 (1.7) 1-1/2 (38)425 (1.9)480 (2.1)400 (1.8)330 (1.5) 365 (1.6) 365 (1.6) Universal Knurled Shank Fasteners X-U*0.157 (4.0) 3/4 (19)125 (0.56)115 (0.51)130 (0.58)95 (0.42) 245 (1.1) 245 (1.1) 1 (25)205 (0.91)260 (1.16) 215 (0.96)155 (0.69) 330 (1.5) 330 (1.5) 1-1/4 (32)315 (1.40)435 (1.93)295 (1.31)200 (0.89) 375 (1.7) 375 (1.7) 1-1/2 (38)425 (1.89)475 (2.11)400 (1.78)260 (1.16) 430 (1.9) 430 (1.9) Standard Fastener X-C (Black collated strip or guidance washer) 0.138 (3.5) 3/4 (19)120 (0.53)175 (0.78)120 (0.53)95 (0.42) 265 (1.2) 265 (1.2) 1 (25)180 (0.80)260 (1.16)215 (0.96)155 (0.69) 485 (2.2) 485 (2.2) 1-1/4 (32)225 (1.00)400 (1.78)250 (1.11)200 (0.89) 500 (2.2) 500 (2.2) 1-1/2 (38)285 (1.27)400 (1.78)285 (1.27)210 (0.93) 555 (2.5) 555 (2.5) Heavy Duty Fastener DS4 0.177 (4.5) 3/4 (19)100 (0.44)200 (0.89)100 (0.44)–200 (0.9) 200 (0.9) 1 (25)180 (0.80)360 (1.60)180 (0.80)180 (0.80) 405 (1.8) 405 (1.8) 1-1/4 (32)300 (1.33)520 (2.31)300 (1.33)250 (1.11) 515 (2.3) 515 (2.3) 1-1/2 (38)450 (2.00)680 (3.02)450 (2.00)325 (1.45) 625 (2.8) 625 (2.8) Stainless Steel Fastener X-CR 0.145 (3.7) 1 (25)230 (1.02)240 (1.07)230 (1.02)–240 (1.1) 240 (1.1) 1-1/4 (32)320 (1.42)400 (1.78)320 (1.42)–400 (1.8) 400 (1.8) 1-1/2 (38)405 (1.80)500 (2.22)405 (1.80)–500 (2.2) 500 (2.2) Gas Fastener X-GN, X-C B3, X-C G3 0.118 (3.0)3/4 (19)115 (0.5)140 (0.6)75 (0.3)85 (0.4) 175 (0.8) 215 (1.0) 1 (25)170 (0.8)220 (1.0)155 (0.7)160 (0.7) 255 (1.1) 315 (1.4) Premium Gas Fastener X-GHP, X-P 17 G2, X-P 20 G2, X-P G3, X-P B3 0.118 (3.0)5/8 (16)60 (0.3)140 (0.6)60 (0.3)60 (0.3) 175 (0.8) 215 (1.0) 1 The tabulated allowable load values are for the low-velocity fasteners only, using a safety factor that is greater than or equal to 5.0, calculated in accordance with ICC- ES AC70. Wood or steel members connected to the substrate must be investigated in accordance with accepted design criteria. 2 The steel deck proՐle is 3 deep composite Ցoor deck with a minimum thickness of 20 gauge (0.0358). Figure 1 (Section 3.2.1.6) shows the nominal Ցute dimensions, fastener locations, and load orientations for the deck proՐle. 3 Structural lightweight concrete Րll above top of metal deck shall be a minimum of 3-1/4" deep. 4 DS fasteners installed at 1-1/2" embedment through steel deck into the lower Ցute must be installed at a minimum distance of 6" from the edge of the Ցoor deck. 5 Multiple fasteners are recommended for any attachment. 1-1/4 (32)325 (1.4)445 (2.0)295 (1.3)230 (1.0)375 (1.7)375 (1.7) www.SSMA.comCopyright © 2015 by the SSMA 6HFWLRQ 'HVLJQ 7KLFNQHVV LQ Fy NVL *URVV3URSHUWLHV (IIHFWLYH3URSHUWLHV 7RUVLRQDO3URSHUWLHV /X LQ Area LQ2 :HLJKW OEIW lx LQ4 Sx LQ3 5[ LQ ,\ LQ4 5\ LQ ,[H LQ4 Sxe LQ3 Mal LQN Mad LQN 9DJ OE 9DQHW OE Jx1000 LQ4 &Z LQ6 Xo LQ P LQ 5R LQ ß 400S250-43 0.0451 33 0.447 1.52 1.224 0.612 1.655 0.399 0.945 1.224 0.503 9.93 10.41 1739 810 0.303 1.486 -2.139 1.252 2.864 0.443 63.7 400S250-54 0.0566 33 0.556 1.89 1.512 0.756 1.649 0.49 0.938 1.512 0.653 12.9 13.91 2603 944 0.594 1.821 -2.124 1.244 2.848 0.444 63.8 400S250-54 0.0566 50 0.556 1.89 1.512 0.756 1.649 0.49 0.938 1.506 0.576 17.24 18.42 3372 1223 0.594 1.821 -2.124 1.244 2.848 0.444 51.6 400S250-68 0.0713 33 0.693 2.36 1.864 0.932 1.64 0.599 0.929 1.864 0.883 17.45 18.42 3215 895 1.174 2.225 -2.105 1.235 2.826 0.445 64 400S250-68 0.0713 50 0.693 2.36 1.864 0.932 1.64 0.599 0.929 1.864 0.775 23.19 24.76 4871 1356 1.174 2.225 -2.105 1.235 2.826 0.445 51.6 400S250-97 0.1017 33 0.966 3.29 2.541 1.271 1.622 0.801 0.911 2.541 1.253 28.31 28.7 4394 797 3.329 2.978 -2.066 1.214 2.78 0.448 60.3 400S250-97 0.1017 50 0.966 3.29 2.541 1.271 1.622 0.801 0.911 2.541 1.191 40.06 41.47 6658 1207 3.329 2.978 -2.066 1.214 2.78 0.448 48.8 400S300-54 0.0566 33 0.613 2.09 1.732 0.866 1.681 0.760 1.114 1.723 0.680 13.44 14.70 2603 944 0.655 2.802 -2.594 1.496 3.285 0.377 74.0 400S300-54 0.0566 50 0.613 2.09 1.732 0.866 1.681 0.760 1.114 1.637 0.592 17.72 19.25 3372 1223 0.655 2.802 -2.594 1.496 3.285 0.377 59.9 400S300-68 0.0713 33 0.764 2.60 2.139 1.070 1.673 0.933 1.105 2.139 0.914 18.06 19.68 3215 895 1.295 3.432 -2.574 1.486 3.263 0.378 74.3 400S300-68 0.0713 50 0.764 2.60 2.139 1.070 1.673 0.933 1.105 2.099 0.805 24.09 26.05 4871 1356 1.295 3.432 -2.574 1.486 3.263 0.378 60.0 400S300-97 0.1017 33 1.067 3.63 2.928 1.464 1.656 1.258 1.086 2.928 1.381 30.58 32.4 4394 797 3.679 4.619 -2.535 1.465 3.216 0.379 70.8 400S300-97 0.1017 50 1.067 3.63 2.928 1.464 1.656 1.258 1.086 2.897 1.307 39.12 40.72 6658 1207 3.679 4.619 -2.535 1.465 3.216 0.379 60.3 550S137-33 0.0346 33 0.301 1.02 1.283 0.467 2.064 0.067 0.472 1.283 0.453 8.95 7.48 699 699 0.12 0.411 -0.841 0.536 2.278 0.864 33.7 550S137-43 0.0451 33 0.391 1.33 1.655 0.602 2.059 0.085 0.467 1.655 0.592 13.08 11.6 1550 1199 0.265 0.52 -0.83 0.53 2.268 0.866 31.7 550S137-54 0.0566 33 0.486 1.65 2.039 0.741 2.049 0.103 0.46 2.039 0.741 16.77 15.9 2739 1666 0.519 0.632 -0.817 0.523 2.254 0.868 31.1 550S137-54 0.0566 50 0.486 1.65 2.039 0.741 2.049 0.103 0.46 2.039 0.714 24.03 20.88 3093 1881 0.519 0.632 -0.817 0.523 2.254 0.868 25.4 550S137-68 0.0713 33 0.604 2.05 2.503 0.91 2.036 0.123 0.451 2.503 0.91 21.22 21.22 4347 2057 1.023 0.764 -0.801 0.514 2.234 0.871 30.4 550S137-68 0.0713 50 0.604 2.05 2.503 0.91 2.036 0.123 0.451 2.503 0.909 31.42 28.89 5350 2532 1.023 0.764 -0.801 0.514 2.234 0.871 24.9 550S137-97 0.1017 33 0.838 2.85 3.38 1.229 2.008 0.155 0.43 3.38 1.229 30.35 30.35 6282 1997 2.891 0.997 -0.766 0.497 2.192 0.878 29.2 550S137-97 0.1017 50 0.838 2.85 3.38 1.229 2.008 0.155 0.43 3.38 1.229 44.72 44.72 9518 3026 2.891 0.997 -0.766 0.497 2.192 0.878 23.9 550S162-33 0.0346 33 0.327 1.11 1.458 0.530 2.112 0.113 0.589 1.458 0.512 10.11 8.63 699 699 0.130 0.713 -1.114 0.697 2.459 0.795 41.4 550S162-43 0.0451 33 0.424 1.44 1.883 0.685 2.107 0.145 0.584 1.883 0.681 14.79 2 13.14 1550 1199 0.288 0.905 -1.103 0.691 2.448 0.797 39.2 550S162-54 0.0566 33 0.528 1.80 2.324 0.845 2.098 0.176 0.577 2.324 0.845 18.76 2 17.87 2739 1666 0.564 1.105 -1.090 0.684 2.434 0.800 38.7 550S162-54 0.0566 50 0.528 1.80 2.324 0.845 2.098 0.176 0.577 2.324 0.811 26.86 2 23.52 3093 1881 0.564 1.105 -1.090 0.684 2.434 0.800 31.6 550S162-68 0.0713 33 0.657 2.24 2.861 1.040 2.086 0.212 0.568 2.861 1.040 23.72 2 23.72 4347 2057 1.114 1.342 -1.072 0.675 2.414 0.803 38.0 550S162-68 0.0713 50 0.657 2.24 2.861 1.040 2.086 0.212 0.568 2.861 1.031 34.94 2 32.28 5350 2532 1.114 1.342 -1.072 0.675 2.414 0.803 31.1 550S162-97 0.1017 33 0.915 3.11 3.886 1.413 2.061 0.276 0.549 3.886 1.413 33.91 33.91 6282 1997 3.154 1.775 -1.037 0.656 2.372 0.809 36.8 550S162-97 0.1017 50 0.915 3.11 3.886 1.413 2.061 0.276 0.549 3.886 1.413 50.13 50.13 9518 3026 3.154 1.775 -1.037 0.656 2.372 0.809 30 550S200-33 0.0346 33 0.362 1.23 1.694 0.616 2.164 0.204 0.751 1.678 0.559 11.05 9.80 699 699 0.144 1.326 -1.508 0.925 2.742 0.698 51.9 550S200-43 0.0451 33 0.469 1.60 2.189 0.796 2.159 0.261 0.746 2.189 0.776 15.33 13.96 1550 1199 0.318 1.691 -1.496 0.918 2.731 0.700 51.7 550S200-54 0.0566 33 0.585 1.99 2.706 0.984 2.152 0.32 0.739 2.706 0.984 21.41 19.98 2739 1666 0.624 2.072 -1.483 0.911 2.716 0.702 49.2 550S200-54 0.0566 50 0.585 1.99 2.706 0.984 2.152 0.32 0.739 2.706 0.901 26.98 24.84 3093 1881 0.624 2.072 -1.483 0.911 2.716 0.702 41.8 550S200-68 0.0713 33 0.729 2.48 3.341 1.215 2.141 0.389 0.731 3.341 1.215 27.03 27.03 4347 2057 1.235 2.531 -1.465 0.902 2.695 0.705 48.5 550S200-68 0.0713 50 0.729 2.48 3.341 1.215 2.141 0.389 0.731 3.341 1.17 38.83 35.92 5350 2532 1.235 2.531 -1.465 0.902 2.695 0.705 39.6 550S200-97 0.1017 33 1.016 3.46 4.563 1.659 2.119 0.515 0.712 4.563 1.659 38.58 38.58 6282 1997 3.504 3.384 -1.428 0.882 2.652 0.710 47.4 550S200-97 0.1017 50 1.016 3.46 4.563 1.659 2.119 0.515 0.712 4.563 1.659 57.25 57.25 9518 3026 3.504 3.384 -1.428 0.882 2.652 0.710 38.6 550S250-43 0.0451 33 0.515 1.75 2.524 0.918 2.215 0.445 0.93 2.524 0.817 16.15 14.74 1550 1199 0.349 2.837 -1.933 1.163 3.083 0.607 62.6 550S250-54 0.0566 33 0.641 2.18 3.126 1.137 2.208 0.547 0.923 3.126 1.033 20.40 19.87 2739 1666 0.685 3.486 -1.919 1.155 3.067 0.609 62.6 550S250-54 0.0566 50 0.641 2.18 3.126 1.137 2.208 0.547 0.923 3.084 0.95 28.44 26.11 3093 1881 0.685 3.486 -1.919 1.155 3.067 0.609 50.7 550S250-68 0.0713 33 0.800 2.72 3.866 1.406 2.198 0.669 0.914 3.866 1.345 29.28 28.52 4347 2057 1.356 4.274 -1.900 1.146 3.046 0.611 59.5 550S250-68 0.0713 50 0.800 2.72 3.866 1.406 2.198 0.669 0.914 3.864 1.233 36.91 35.43 5350 2532 1.356 4.274 -1.900 1.146 3.046 0.611 50.6 550S250-97 0.1017 33 1.118 3.80 5.304 1.929 2.178 0.897 0.895 5.304 1.925 43.47 43.57 6282 1997 3.855 5.761 -1.862 1.126 3.002 0.615 58.4 550S250-97 0.1017 50 1.118 3.80 5.304 1.929 2.178 0.897 0.895 5.304 1.837 61.77 60.32 9518 3026 3.855 5.761 -1.862 1.126 3.002 0.615 47.6 600S137-33 0.0346 33 0.318 1.08 1.582 0.527 2.229 0.069 0.464 1.548 0.455 8.98 8.19 638 638 0.127 0.500 -0.807 0.519 2.416 0.889 33.5 600S137-43 0.0451 33 0.413 1.41 2.042 0.681 2.223 0.087 0.459 2.041 0.645 12.74 11.82 1416 1240 0.280 0.633 -0.796 0.513 2.406 0.890 33.3 600S137-54 0.0566 33 0.514 1.75 2.518 0.839 2.213 0.105 0.452 2.518 0.832 16.44 15.95 2739 1890 0.549 0.769 -0.784 0.506 2.391 0.893 33.0 600S137-54 0.0566 50 0.514 1.75 2.518 0.839 2.213 0.105 0.452 2.518 0.777 23.26 21.24 2823 1947 0.549 0.769 -0.784 0.506 2.391 0.893 26.8 600S137-68 0.0713 33 0.640 2.18 3.094 1.031 2.200 0.125 0.443 3.094 1.031 24.05 2 24.05 4347 2339 1.084 0.930 -0.768 0.497 2.371 0.895 30.1 600S137-68 0.0713 50 0.640 2.18 3.094 1.031 2.200 0.125 0.443 3.094 1.030 30.84 28.89 5350 2879 1.084 0.930 -0.768 0.497 2.371 0.895 26.5 600S137-97 0.1017 33 0.889 3.03 4.188 1.396 2.170 0.159 0.422 4.188 1.396 34.48 2 34.49 6911 2512 3.066 1.216 -0.734 0.480 2.330 0.901 28.8 600S137-97 0.1017 50 0.889 3.03 4.188 1.396 2.170 0.159 0.422 4.188 1.396 50.80 2 50.80 10472 3805 3.066 1.216 -0.734 0.480 2.330 0.901 23.6 600S137-118 0.1242 33 1.065 3.62 4.913 1.638 2.147 0.176 0.406 4.913 1.638 42.05 42.05 8267 2391 5.477 1.391 -0.709 0.467 2.298 0.905 27.9 600S137-118 0.1242 50 1.065 3.62 4.913 1.638 2.147 0.176 0.406 4.913 1.638 61.69 61.69 12526 3622 5.477 1.391 -0.709 0.467 2.298 0.905 22.9 600S162-33 0.0346 33 0.344 1.17 1.793 0.598 2.282 0.116 0.581 1.793 0.577 11.41 9.47 638 638 0.137 0.861 -1.072 0.677 2.587 0.828 41.1 600S162-43 0.0451 33 0.447 1.52 2.316 0.772 2.276 0.148 0.576 2.316 0.767 16.68 2 14.46 1416 1240 0.303 1.095 -1.062 0.670 2.577 0.830 39.0 600S162-54 0.0566 33 0.556 1.89 2.860 0.953 2.267 0.180 0.570 2.860 0.953 21.17 2 19.75 2739 1890 0.594 1.337 -1.049 0.663 2.562 0.832 38.4 600S162-54 0.0566 50 0.556 1.89 2.860 0.953 2.267 0.180 0.570 2.860 0.916 30.33 2 25.90 2823 1947 0.594 1.337 -1.049 0.663 2.562 0.832 31.4 600S162-68 0.0713 33 0.693 2.36 3.525 1.175 2.255 0.218 0.560 3.525 1.175 26.79 2 26.78 4347 2339 1.174 1.626 -1.032 0.655 2.543 0.835 37.7 600S162-68 0.0713 50 0.693 2.36 3.525 1.175 2.255 0.218 0.560 3.525 1.164 39.47 2 35.69 5350 2879 1.174 1.626 -1.032 0.655 2.543 0.835 30.8 600S162-97 0.1017 33 0.966 3.29 4.797 1.599 2.229 0.283 0.541 4.797 1.599 38.37 2 38.37 6911 2512 3.329 2.153 -0.997 0.636 2.501 0.841 36.4 600S162-97 0.1017 50 0.966 3.29 4.797 1.599 2.229 0.283 0.541 4.797 1.599 56.73 2 56.72 10472 3805 3.329 2.153 -0.997 0.636 2.501 0.841 29.8 600S162-118 0.1242 33 1.158 3.94 5.652 1.884 2.209 0.321 0.526 5.652 1.884 46.82 2 46.82 8267 2391 5.956 2.487 -0.971 0.623 2.470 0.845 35.6 600S162-118 0.1242 50 1.158 3.94 5.652 1.884 2.209 0.321 0.526 5.652 1.884 68.94 2 68.93 12526 3622 5.956 2.487 -0.971 0.623 2.470 0.845 29.1 1Web height-to-thickness ratio exceeds 200. Web stiffeners are required at all support points and concentrated loads. 2Allowable moment includes cold work of forming. See Table Notes on page 7. Structural (S) Section Properties 600S162-43 0.0451 33 0.447 1.52 2.316 0.772 2.276 0.14 8 0.576 2.316 0.767 16.68 14.46 1416 1240 0.303 1.095 -1.062 0.670 2.577 0.830 39.02 Project Job Ref. Section Existing Column-Beam Frame Sheet no./rev. 1 Calc. by N Date 3/27/2020 Chk'd by Date App'd by Date STEEL MEMBER ANALYSIS & DESIGN (AISC 360) In accordance with AISC360 14th Edition published 2010 using the LRFD method Tedds calculation version 4.3.04 ANALYSIS Tedds calculation version 1.0.27 Geometry Geometry (ft) - Steel (AISC) M e m b e r 1 M e m b e r 2 Member3 Member4 M e m b e r 5 M e m b e r 6 Member7 16 2 5 . 8 3 1 35.831 4 6 5 20.5 6 6 76 8 5 . 8 3 1 95.831 1 2 3 4 5 6 7 8 X Z Materials Name Density Youngs Modulus Shear Modulus Thermal Coefficient (lbm/ft3) ksi ksi C-1 Steel (AISC) 490 29000 11200 0.000012 Sections Name Area Moment of inertia Shear area parallel to Major Minor Minor Major (in2) (in 4) (in 4) (in 2) (in 2) W 18x50 15 800 40 6 8 HSS 12x8x1/4 9 184 99 5 4 Nodes Node Co-ordinates Freedom Coordinate system Spring X Z X Z Rot. Name Angle X Z Rot. (ft) (ft)()(kips/ft) (kips/ft)kip_ft/ 1 3 0 Fixed Fixed Fixed 0 0 0 0 2 3 6 Fixed Fixed Free 0 0 0 0 3 0 11 Free Free Free 0 0 0 0 4 6 11 Free Free Free 0 0 0 0 5 26.5 11 Free Free Free 0 0 0 0 6 32.5 11 Free Free Free 0 0 0 0 7 29.5 6 Fixed Fixed Free 0 0 0 0 8 29.5 0 Fixed Fixed Fixed 0 0 0 0 Project Job Ref. Section Existing Column-Beam Frame Sheet no./rev. 2 Calc. by N Date 3/27/2020 Chk'd by Date App'd by Date Elements Element Length Nodes Section Material Releases Rotated (ft) Start End Start moment End moment Axial 1 6 1 2 HSS 12x8x1/4 Steel (AISC) Fixed Fixed Fixed 2 5.83 2 3 HSS 12x8x1/4 Steel (AISC) Fixed Fixed Fixed 3 5.83 2 4 HSS 12x8x1/4 Steel (AISC) Fixed Fixed Fixed 4 6 3 4 W 18x50 Steel (AISC) Fixed Fixed Fixed 5 20.5 4 5 W 18x50 Steel (AISC) Fixed Fixed Fixed 6 6 5 6 W 18x50 Steel (AISC) Fixed Fixed Fixed 7 6 8 7 HSS 12x8x1/4 Steel (AISC) Fixed Fixed Fixed 8 5.83 7 5 HSS 12x8x1/4 Steel (AISC) Fixed Fixed Fixed 9 5.83 7 6 HSS 12x8x1/4 Steel (AISC) Fixed Fixed Fixed Members Name Elements Start End Member1 1 1 Member2 2 2 Member3 3 3 Member4 4 6 Member5 7 7 Member6 8 8 Member7 9 9 Loading Self weight included Dead - Loading (kips/ft,kips) 9 . 9 1. 6 1. 6 1. 6 1. 6 9 . 9 1. 6 1. 6 Me m b e r 1 M e m b e r 2 Member3 Member4 Me m b e r 5 M e m b e r 6 Member7 X Z Project Job Ref. Section Existing Column-Beam Frame Sheet no./rev. 3 Calc. by N Date 3/27/2020 Chk'd by Date App'd by Date Live - Loading (kips/ft) 2. 3 2. 3 2. 3 2. 3 2. 3 2. 3 Me m b e r 1 M e m b e r 2 Member3 Member4 Me m b e r 5 M e m b e r 6 Member7 X Z Snow - Loading (kips/ft,kips) 2 7 2 7 Me m b e r 1 M e m b e r 2 Member3 Member4 Me m b e r 5 M e m b e r 6 Member7 X Z Wind - Loading (kips) 7 Me m b e r 1 M e m b e r 2 Member3 Member4 Me m b e r 5 M e m b e r 6 Member7 X Z Load combination factors Load combination Se l f W e i g h t De a d Li v e Sn o w Wi n d 1.2D + 1.6L (Strength) 1.20 1.20 1.60 Project Job Ref. Section Existing Column-Beam Frame Sheet no./rev. 4 Calc. by N Date 3/27/2020 Chk'd by Date App'd by Date Load combination Se l f W e i g h t De a d Li v e Sn o w Wi n d 1.0D + 1.0L (Service) 1.00 1.00 1.00 1.2D + 1.6L + 0.5S (Strength) 1.20 1.20 1.60 0.50 1.0D + 0.75L + 0.75S (Service) 1.00 1.00 0.75 0.75 1.2D + 1.0L + 1.6S (Strength) 1.20 1.20 1.00 1.60 1.0D + 0.0L + 1.0S (Service) 1.00 1.00 0.00 1.00 1.2D + 1.6S + 0.5W (Strength) 1.20 1.20 1.60 0.50 1.0D + 1.0S + 0.5W (Service) 1.00 1.00 1.00 0.50 1.2D + 1.0L + 0.5S + 1.0W (Strength) 1.20 1.20 1.00 0.50 1.00 1.0D + 0.5L + 0.5S + 0.7W (Service) 1.00 1.00 0.50 0.50 0.70 0.9D + 1.0W (Strength) 0.90 0.90 1.00 Node loads Node Load case Force Moment X Z (kips) (kips) (kip_ft) 3 Wind 7 0 0 Member Loads Member Load case Load Type Orientation Description Member4 Dead UDL GlobalZ 1.6 kips/ft Member4 Live UDL GlobalZ 2.3 kips/ft Element Loads Element Load case Load Type Orientation Description 4 Dead UDL GlobalZ 0 kips/ft 4 Dead Point load GlobalZ 9.9 kips at 3 ft 6 Dead Point load GlobalZ 9.9 kips at 3 ft 4 Snow UDL GlobalZ 0 kips/ft 4 Snow Point load GlobalZ 27 kips at 3 ft 6 Snow Point load GlobalZ 27 kips at 3 ft Project Job Ref. Section Existing Column-Beam Frame Sheet no./rev. 5 Calc. by N Date 3/27/2020 Chk'd by Date App'd by Date Results Forces Strength combinations - Moment envelope (kip_ft) 2 -4.7 16.1 -4.7 2.6 -31.9 54.6 -134.8 135.3 -163.7 50.8 -134.8 1.7 -4 2.6 -31.9 16.1 -6.4 Strength combinations - Shear envelope (kips) 0 -1.1 3.5 0.1 1.5 -6 26 -56.7 58 -58 56.7 -24.9 0 -0.9 0.5 -6 3.7 0.5 Service combinations - Deflection envelope (in) 0 0 0 0 0 0 0 0 0.3 0 0 0 0 0 0 0 0 ; Resistance factors Shear;v = 0.90 Flexure;b = 0.90 Project Job Ref. Section Existing Column-Beam Frame Sheet no./rev. 6 Calc. by N Date 3/27/2020 Chk'd by Date App'd by Date Tensile yielding;t,y = 0.90 Tensile rupture;t,r = 0.75 Compression;c = 0.90 Member1 design Section details Section type;HSS 12x8x1/4 (AISC 15th Edn (v15.0)) ASTM steel designation;User defined Steel yield stress;Fy = 42 ksi Steel tensile stress;Fu = 65 ksi Modulus of elasticity;E = 29000 ksi 8" 0.23" 12 " HSS 12x8x1/4 (AISC 15th Edn (v15.0)) Section depth, d, 12 in Section breadth, bf, 8 in Weight of section, Weight, 32.6 lbf/ft Section thickness, t, 0.233 in Area of section, A, 9 in2 Radius of gyration about x-axis, r x, 4.53 in Radius of gyration about y-axis, r y, 3.32 in Elastic section modulus about x-axis, Sx, 30.6 in3 Elastic section modulus about y-axis, Sy, 24.7 in3 Plastic section modulus about x-axis, Zx, 36.6 in3 Plastic section modulus about y-axis, Zy, 27.8 in3 Second moment of area about x-axis, I x, 184 in4 Second moment of area about y-axis, I y, 98.8 in4 Lateral restraint Both flanges have lateral restraint at supports only Consider Combination 9 - 1.2D + 1.0L + 0.5S + 1.0W (Strength) Classification of sections for local buckling - Section B4 Classification of flanges in flexure - Table B4.1b (case 17) Width to thickness ratio;(bf - 3  t) / t = 31.33 Limiting ratio for compact section;pff = 1.12  [E / Fy] = 29.43 Limiting ratio for non-compact section;rff = 1.40  [E / Fy] = 36.79;Noncompact Classification of web in flexure - Table B4.1b (case 19) Width to thickness ratio;(d - 3  t) / t = 48.50 Limiting ratio for compact section;pwf = 2.42  [E / Fy] = 63.59 Limiting ratio for non-compact section;rwf = 5.70  [E / Fy] = 149.78;Compact Section is noncompact in flexure Classification of flanges in uniform compression - Table B4.1a (case 6) Width to thickness ratio;(bf - 3  t) / t = 31.33 Limiting ratio for non-compact section;rfc = 1.40  [E / Fy] = 36.79;Nonslender Project Job Ref. Section Existing Column-Beam Frame Sheet no./rev. 7 Calc. by N Date 3/27/2020 Chk'd by Date App'd by Date Classification of web in uniform compression - Table B4.1a (case 6) Width to thickness ratio;(d - 3  t) / t = 48.50 Limiting ratio for non-compact section;rwc = 1.40  [E / Fy] = 36.79;Slender Section is slender in compression Check design at start of span Design of members for compression - Chapter E Required compressive strength;Pr = "Expresso error (169)" kips Check design at end of span Design of members for tension - Chapter D Required tensile strength;Pr = 0.11 kips Slenderness limitations - Section D1 Slenderness ratio; = max(Lb,x / rx, Lb,y / ry) = ?<Error: invalid expression> Tension member slenderness ratio does not exceed recommended limit of 300 Design tensile strength - D2 Nominal tensile yielding strength;Pn,yld = Fy  A = 376.32 kips Nominal tensile rupture strength;Pn,r = Fu  Ae = 522.855 kips Design tensile strength;Pc = min(t,y  Pn,yld, t,r  Pn,r) = 338.688 kips Pr / Pc = 0 PASS - Nominal tensile strength exceeds required tensile strength Design of members for shear - Chapter G Required shear strength;Vr,x = 1.1 kips Web area;Aw = 2  (d - 3  t)  t = 5.266 in2 Web plate buckling coefficient;kv = 5 (d - 3  t) / t <= 1.10  (kv  E / Fy) Web shear coefficient - eq G2-3;Cv = 1.000 Nominal shear strength - eq G2-1;Vn,x = 0.6  Fy  Aw  Cv = 132.7 kips Resistance factor;v = 0.90 Design shear strength;Vc,x = v  Vn,x = 119.4 kips Vr,x / Vc,x = 0.009 PASS - Design shear strength exceeds required shear strength Design of members for flexure - Chapter F Required flexural strength;Mr,x = 4.7 kips_ft Yielding - Section F7.1 Nominal flexural strength for yielding - eq F7-1;Mn,yld,x = Mp,x = Fy  Zx = 128.1 kips_ft Compression flange local buckling - Section F7.2 Nominal flexural strength for compression flange local buckling - eq F7-2 ; Mn,flb,x = min(Mp,x - (Mp,x - Fy  Sx)  (3.57  (bf - 3  t) / t  (Fy / E) - 4.0), Mp,x) = 122.7 kips_ft Web local buckling - Section F7.3 Nominal flexural strength for web local buckling - eq F7-5; Mn,wlb,x = Mp,x = 128.1 kips_ft Project Job Ref. Section Existing Column-Beam Frame Sheet no./rev. 8 Calc. by N Date 3/27/2020 Chk'd by Date App'd by Date Design flexural strength - F1 Nominal flexural strength;Mn,x = min(Mn,yld,x, Mn,flb,x) = 122.7 kips_ft Design flexural strength;Mc,x = b  Mn,x = 110.4 kips_ft Mr,x / Mc,x = 0.043 PASS - Design flexural strength exceeds required flexural strength Design of members for combined forces - Chapter H Combined flexure and axial force - eq H1-1b;Pr / (2  Pc) + Mr,x / Mc,x = 0.043 PASS - Combined flexure and axial force is within acceptable limits Consider Combination 10 - 1.0D + 0.5L + 0.5S + 0.7W (Service) Check design 3ft 10.859in along span Design of members for x-x axis deflection Maximum deflection;x = 0.002 in Allowable deflection;x,Allowable = Lm1_s1 / 360 = 0.2 in x / x,Allowable = 0.008 PASS - Allowable deflection exceeds design deflection Member2 design Section details Section type;HSS 12x8x1/4 (AISC 15th Edn (v15.0)) ASTM steel designation;User defined Steel yield stress;Fy = 42 ksi Steel tensile stress;Fu = 65 ksi Modulus of elasticity;E = 29000 ksi 8" 0.23" 1 2 " HSS 12x8x1/4 (AISC 15th Edn (v15.0)) Section depth, d, 12 in Section breadth, bf, 8 in Weight of section, Weight, 32.6 lbf/ft Section thickness, t, 0.233 in Area of section, A, 9 in2 Radius of gyration about x-axis, r x, 4.53 in Radius of gyration about y-axis, r y, 3.32 in Elastic section modulus about x-axis, Sx, 30.6 in3 Elastic section modulus about y-axis, Sy, 24.7 in3 Plastic section modulus about x-axis, Zx, 36.6 in3 Plastic section modulus about y-axis, Zy, 27.8 in3 Second moment of area about x-axis, I x, 184 in4 Second moment of area about y-axis, I y, 98.8 in4 Lateral restraint Both flanges have lateral restraint at supports only Consider Combination 5 - 1.2D + 1.0L + 1.6S (Strength) Classification of sections for local buckling - Section B4 Project Job Ref. Section Existing Column-Beam Frame Sheet no./rev. 9 Calc. by N Date 3/27/2020 Chk'd by Date App'd by Date Classification of flanges in flexure - Table B4.1b (case 17) Width to thickness ratio;(bf - 3  t) / t = 31.33 Limiting ratio for compact section;pff = 1.12  [E / Fy] = 29.43 Limiting ratio for non-compact section;rff = 1.40  [E / Fy] = 36.79;Noncompact Classification of web in flexure - Table B4.1b (case 19) Width to thickness ratio;(d - 3  t) / t = 48.50 Limiting ratio for compact section;pwf = 2.42  [E / Fy] = 63.59 Limiting ratio for non-compact section;rwf = 5.70  [E / Fy] = 149.78;Compact Section is noncompact in flexure Classification of flanges in uniform compression - Table B4.1a (case 6) Width to thickness ratio;(bf - 3  t) / t = 31.33 Limiting ratio for non-compact section;rfc = 1.40  [E / Fy] = 36.79;Nonslender Classification of web in uniform compression - Table B4.1a (case 6) Width to thickness ratio;(d - 3  t) / t = 48.50 Limiting ratio for non-compact section;rwc = 1.40  [E / Fy] = 36.79;Slender Section is slender in compression Check design at start of span Design of members for compression - Chapter E Required compressive strength;Pr = "Expresso error (169)" kips Check design at end of span Design of members for compression - Chapter E Required compressive strength;Pr = 26 kips Slenderness limitations and effective length - Section E2 Unbraced length;Lb,x = Lm2_s1 = 5.831 ft Effective length factor;Kx = 1.00 Column slenderness;x = Kx  Lb,x / rx = 15.446 Major axis column slenderness ratio does not exceed recommended limit of 200 Slenderness limitations and effective length - Section E2 Unbraced length;Lb,y = Lm2_s1_seg1 = 5.831 ft Effective length factor;Ky = 1.00 Column slenderness;y = Ky  Lb,y / ry = 21.076 Minor axis column slenderness ratio does not exceed recommended limit of 200 Flexural buckling of members with slender elements - Section E7 Width of stiffened element b = bf - 3 t = 7.301 in Reduced effective width - eq E7-18 be = b = 7.3 in Height of stiffened compression element h = d - 3 t = 11.301 in Reduced effective height - eq E7-18 he = min(1.92  t  (E / f)  [1 - (0.38 / (h / t))  (E / f)], h) = 9.3 in Effective area of cross-section Ae = A - 2  (h - he)  t = 8.044 in2 Slender stiffened element factor - eq E7-16 Qa = Ae / A = 0.898 Net reduction factor;Q = Qa = 0.898 Elastic critical buckling stress - eq E3-4;Fe,x = 2  E / (Kx  Lb,x / rx)2 = 1199.6 ksi Project Job Ref. Section Existing Column-Beam Frame Sheet no./rev. 10 Calc. by N Date 3/27/2020 Chk'd by Date App'd by Date Flexural buckling stress - eq E7-2;Fcr,x = Q  [0.658Q  Fy / Fe,x]  Fy = 37.2 ksi Nominal compressive strength for flexural buckling - eq E7-1; Pn,fb,x = Fcr,x  A = 333.4 kips Width of stiffened element b = bf - 3 t = 7.301 in Reduced effective width - eq E7-18 be = b = 7.3 in Height of stiffened compression element h = d - 3 t = 11.301 in Reduced effective height - eq E7-18 he = min(1.92  t  (E / f)  [1 - (0.38 / (h / t))  (E / f)], h) = 9.3 in Effective area of cross-section Ae = A - 2  (h - he)  t = 8.044 in2 Slender stiffened element factor - eq E7-16 Qa = Ae / A = 0.898 Net reduction factor;Q = Qa = 0.898 Elastic critical buckling stress - eq E3-4;Fe,y = 2  E / (Ky  Lb,y / ry)2 = 644.4 ksi Flexural buckling stress - eq E7-2;Fcr,y = Q  [0.658Q  Fy / Fe,y]  Fy = 36.8 ksi Nominal compressive strength for flexural buckling - eq E7-1; Pn,fb,y = Fcr,y  A = 329.7 kips Design compressive strength - E1 Nominal compressive strength;Pn = min(Pn,fb,x, Pn,fb,y) = 329.7 kips Design compressive strength;Pc = c  Pn = 296.7 kips Pr / Pc = 0.088 PASS - Nominal compressive strength exceeds required compressive strength Design of members for shear - Chapter G Required shear strength;Vr,x = 3.4 kips Web area;Aw = 2  (d - 3  t)  t = 5.266 in2 Web plate buckling coefficient;kv = 5 (d - 3  t) / t <= 1.10  (kv  E / Fy) Web shear coefficient - eq G2-3;Cv = 1.000 Nominal shear strength - eq G2-1;Vn,x = 0.6  Fy  Aw  Cv = 132.7 kips Resistance factor;v = 0.90 Design shear strength;Vc,x = v  Vn,x = 119.4 kips Vr,x / Vc,x = 0.028 PASS - Design shear strength exceeds required shear strength Design of members for flexure - Chapter F Required flexural strength;Mr,x = 16.1 kips_ft Yielding - Section F7.1 Nominal flexural strength for yielding - eq F7-1;Mn,yld,x = Mp,x = Fy  Zx = 128.1 kips_ft Compression flange local buckling - Section F7.2 Nominal flexural strength for compression flange local buckling - eq F7-2 ; Mn,flb,x = min(Mp,x - (Mp,x - Fy  Sx)  (3.57  (bf - 3  t) / t  (Fy / E) - 4.0), Mp,x) = 122.7 kips_ft Web local buckling - Section F7.3 Nominal flexural strength for web local buckling - eq F7-5; Mn,wlb,x = Mp,x = 128.1 kips_ft Project Job Ref. Section Existing Column-Beam Frame Sheet no./rev. 11 Calc. by N Date 3/27/2020 Chk'd by Date App'd by Date Design flexural strength - F1 Nominal flexural strength;Mn,x = min(Mn,yld,x, Mn,flb,x) = 122.7 kips_ft Design flexural strength;Mc,x = b  Mn,x = 110.4 kips_ft Mr,x / Mc,x = 0.146 PASS - Design flexural strength exceeds required flexural strength Design of members for combined forces - Chapter H Combined flexure and axial force - eq H1-1b;Pr / (2  Pc) + Mr,x / Mc,x = 0.190 PASS - Combined flexure and axial force is within acceptable limits Consider Combination 10 - 1.0D + 0.5L + 0.5S + 0.7W (Service) Check design at end of span Design of members for x-x axis deflection Maximum deflection;x = 0.024 in Allowable deflection;x,Allowable = Lm2_s1 / 360 = 0.194 in x / x,Allowable = 0.123 PASS - Allowable deflection exceeds design deflection Member3 design Section details Section type;HSS 12x8x1/4 (AISC 15th Edn (v15.0)) ASTM steel designation;User defined Steel yield stress;Fy = 42 ksi Steel tensile stress;Fu = 65 ksi Modulus of elasticity;E = 29000 ksi 8" 0.23" 1 2 " HSS 12x8x1/4 (AISC 15th Edn (v15.0)) Section depth, d, 12 in Section breadth, bf, 8 in Weight of section, Weight, 32.6 lbf/ft Section thickness, t, 0.233 in Area of section, A, 9 in2 Radius of gyration about x-axis, r x, 4.53 in Radius of gyration about y-axis, r y, 3.32 in Elastic section modulus about x-axis, Sx, 30.6 in3 Elastic section modulus about y-axis, Sy, 24.7 in3 Plastic section modulus about x-axis, Zx, 36.6 in3 Plastic section modulus about y-axis, Zy, 27.8 in3 Second moment of area about x-axis, I x, 184 in4 Second moment of area about y-axis, I y, 98.8 in4 Lateral restraint Both flanges have lateral restraint at supports only Classification of sections for local buckling - Section B4 Classification of flanges in flexure - Table B4.1b (case 17) Width to thickness ratio;(bf - 3  t) / t = 31.33 Project Job Ref. Section Existing Column-Beam Frame Sheet no./rev. 12 Calc. by N Date 3/27/2020 Chk'd by Date App'd by Date Limiting ratio for compact section;pff = 1.12  [E / Fy] = 29.43 Limiting ratio for non-compact section;rff = 1.40  [E / Fy] = 36.79;Noncompact Classification of web in flexure - Table B4.1b (case 19) Width to thickness ratio;(d - 3  t) / t = 48.50 Limiting ratio for compact section;pwf = 2.42  [E / Fy] = 63.59 Limiting ratio for non-compact section;rwf = 5.70  [E / Fy] = 149.78;Compact Section is noncompact in flexure Classification of flanges in uniform compression - Table B4.1a (case 6) Width to thickness ratio;(bf - 3  t) / t = 31.33 Limiting ratio for non-compact section;rfc = 1.40  [E / Fy] = 36.79;Nonslender Classification of web in uniform compression - Table B4.1a (case 6) Width to thickness ratio;(d - 3  t) / t = 48.50 Limiting ratio for non-compact section;rwc = 1.40  [E / Fy] = 36.79;Slender Section is slender in compression Check design at start of span Design of members for compression - Chapter E Required compressive strength;Pr = "Expresso error (169)" kips Check design at end of span Design of members for compression - Chapter E Required compressive strength;Pr = 129.7 kips Slenderness limitations and effective length - Section E2 Unbraced length;Lb,x = Lm3_s1 = 5.831 ft Effective length factor;Kx = 1.00 Column slenderness;x = Kx  Lb,x / rx = 15.446 Major axis column slenderness ratio does not exceed recommended limit of 200 Slenderness limitations and effective length - Section E2 Unbraced length;Lb,y = Lm3_s1_seg1 = 5.831 ft Effective length factor;Ky = 1.00 Column slenderness;y = Ky  Lb,y / ry = 21.076 Minor axis column slenderness ratio does not exceed recommended limit of 200 Flexural buckling of members with slender elements - Section E7 Width of stiffened element b = bf - 3 t = 7.301 in Reduced effective width - eq E7-18 be = b = 7.3 in Height of stiffened compression element h = d - 3 t = 11.301 in Reduced effective height - eq E7-18 he = min(1.92  t  (E / f)  [1 - (0.38 / (h / t))  (E / f)], h) = 9.3 in Effective area of cross-section Ae = A - 2  (h - he)  t = 8.044 in2 Slender stiffened element factor - eq E7-16 Qa = Ae / A = 0.898 Net reduction factor;Q = Qa = 0.898 Elastic critical buckling stress - eq E3-4;Fe,x = 2  E / (Kx  Lb,x / rx)2 = 1199.6 ksi Flexural buckling stress - eq E7-2;Fcr,x = Q  [0.658Q  Fy / Fe,x]  Fy = 37.2 ksi Nominal compressive strength for flexural buckling - eq E7-1; Project Job Ref. Section Existing Column-Beam Frame Sheet no./rev. 13 Calc. by N Date 3/27/2020 Chk'd by Date App'd by Date Pn,fb,x = Fcr,x  A = 333.4 kips Width of stiffened element b = bf - 3 t = 7.301 in Reduced effective width - eq E7-18 be = b = 7.3 in Height of stiffened compression element h = d - 3 t = 11.301 in Reduced effective height - eq E7-18 he = min(1.92  t  (E / f)  [1 - (0.38 / (h / t))  (E / f)], h) = 9.3 in Effective area of cross-section Ae = A - 2  (h - he)  t = 8.044 in2 Slender stiffened element factor - eq E7-16 Qa = Ae / A = 0.898 Net reduction factor;Q = Qa = 0.898 Elastic critical buckling stress - eq E3-4;Fe,y = 2  E / (Ky  Lb,y / ry)2 = 644.4 ksi Flexural buckling stress - eq E7-2;Fcr,y = Q  [0.658Q  Fy / Fe,y]  Fy = 36.8 ksi Nominal compressive strength for flexural buckling - eq E7-1; Pn,fb,y = Fcr,y  A = 329.7 kips Design compressive strength - E1 Nominal compressive strength;Pn = min(Pn,fb,x, Pn,fb,y) = 329.7 kips Design compressive strength;Pc = c  Pn = 296.7 kips Pr / Pc = 0.437 PASS - Nominal compressive strength exceeds required compressive strength Design of members for shear - Chapter G Required shear strength;Vr,x = 5.2 kips Web area;Aw = 2  (d - 3  t)  t = 5.266 in2 Web plate buckling coefficient;kv = 5 (d - 3  t) / t <= 1.10  (kv  E / Fy) Web shear coefficient - eq G2-3;Cv = 1.000 Nominal shear strength - eq G2-1;Vn,x = 0.6  Fy  Aw  Cv = 132.7 kips Resistance factor;v = 0.90 Design shear strength;Vc,x = v  Vn,x = 119.4 kips Vr,x / Vc,x = 0.043 PASS - Design shear strength exceeds required shear strength Design of members for flexure - Chapter F Required flexural strength;Mr,x = 28.9 kips_ft Yielding - Section F7.1 Nominal flexural strength for yielding - eq F7-1;Mn,yld,x = Mp,x = Fy  Zx = 128.1 kips_ft Compression flange local buckling - Section F7.2 Nominal flexural strength for compression flange local buckling - eq F7-2 ; Mn,flb,x = min(Mp,x - (Mp,x - Fy  Sx)  (3.57  (bf - 3  t) / t  (Fy / E) - 4.0), Mp,x) = 122.7 kips_ft Web local buckling - Section F7.3 Nominal flexural strength for web local buckling - eq F7-5; Mn,wlb,x = Mp,x = 128.1 kips_ft Design flexural strength - F1 Nominal flexural strength;Mn,x = min(Mn,yld,x, Mn,flb,x) = 122.7 kips_ft Design flexural strength;Mc,x = b  Mn,x = 110.4 kips_ft Project Job Ref. Section Existing Column-Beam Frame Sheet no./rev. 14 Calc. by N Date 3/27/2020 Chk'd by Date App'd by Date Mr,x / Mc,x = 0.262 PASS - Design flexural strength exceeds required flexural strength Design of members for combined forces - Chapter H Combined flexure and axial force - eq H1-1a;Pr / Pc + 8 / 9  (Mr,x / Mc,x) = 0.670 PASS - Combined flexure and axial force is within acceptable limits Consider Combination 10 - 1.0D + 0.5L + 0.5S + 0.7W (Service) Check design at end of span Design of members for x-x axis deflection Maximum deflection;x = 0.037 in Allowable deflection;x,Allowable = Lm3_s1 / 360 = 0.194 in x / x,Allowable = 0.189 PASS - Allowable deflection exceeds design deflection Member4 - Span 1 design Section details Section type;W 18x50 (AISC 15th Edn (v15.0)) ASTM steel designation;A992 Steel yield stress;Fy = 50 ksi Steel tensile stress;Fu = 65 ksi Modulus of elasticity;E = 29000 ksi 7.5" 0.36" 18 " 0. 5 7 " 0 . 5 7 " W 18x50 (AISC 15th Edn (v15.0)) Section depth, d, 18 in Section breadth, bf, 7.5 in Weight of section, Weight, 50 lbf/ft Flange thickness, tf, 0.57 in Web thickness, tw, 0.355 in Area of section, A, 14.7 in2 Radius of gyration about x-axis, rx, 7.38 in Radius of gyration about y-axis, ry, 1.65 in Elastic section modulus about x-axis, Sx, 88.9 in3 Elastic section modulus about y-axis, Sy, 10.7 in3 Plastic section modulus about x-axis, Zx, 101 in3 Plastic section modulus about y-axis, Zy, 16.6 in3 Second moment of area about x-axis, I x, 800 in4 Second moment of area about y-axis, I y, 40.1 in4 Lateral restraint Both flanges have lateral restraint at supports only Classification of sections for local buckling - Section B4 Classification of flanges in flexure - Table B4.1b (case 10) Width to thickness ratio;bf / (2  tf) = 6.58 Limiting ratio for compact section;pff = 0.38  [E / Fy] = 9.15 Limiting ratio for non-compact section;rff = 1.0  [E / Fy] = 24.08;Compact Project Job Ref. Section Existing Column-Beam Frame Sheet no./rev. 15 Calc. by N Date 3/27/2020 Chk'd by Date App'd by Date Classification of web in flexure - Table B4.1b (case 15) Width to thickness ratio;(d - 2  k) / tw = 45.23 Limiting ratio for compact section;pwf = 3.76  [E / Fy] = 90.55 Limiting ratio for non-compact section;rwf = 5.70  [E / Fy] = 137.27;Compact Section is compact in flexure Classification of flanges in uniform compression - Table B4.1a (case 1) Width to thickness ratio;bf / (2  tf) = 6.58 Limiting ratio for non-compact section;rfc = 0.56  [E / Fy] = 13.49;Nonslender Classification of web in uniform compression - Table B4.1a (case 5) Width to thickness ratio;(d - 2  k) / tw = 45.23 Limiting ratio for non-compact section;rwc = 1.49  [E / Fy] = 35.88;Slender Section is slender in compression Check design at end of span Design of members for tension - Chapter D Required tensile strength;Pr = 3.647 kips Slenderness limitations - Section D1 Slenderness ratio; = max(Lb,x / rx, Lb,y / ry) = 42.407 Tension member slenderness ratio does not exceed recommended limit of 300 Design tensile strength - D2 Nominal tensile yielding strength;Pn,yld = Fy  A = 735 kips Nominal tensile rupture strength;Pn,r = Fu  Ae = 522.855 kips Design tensile strength;Pc = min(t,y  Pn,yld, t,r  Pn,r) = 392.141 kips Pr / Pc = 0.009 PASS - Nominal tensile strength exceeds required tensile strength Design of members for shear - Chapter G Required shear strength;Vr,x = 50.5 kips Web area;Aw = d  tw = 6.39 in2 Web plate buckling coefficient;kv = 5 (d - 2  k) / tw <= 2.24  (E / Fy) Web shear coefficient - eq G2-2;Cv = 1.000 Nominal shear strength - eq G2-1;Vn,x = 0.6  Fy  Aw  Cv = 191.7 kips Resistance factor;v = 1.00 Design shear strength;Vc,x = v  Vn,x = 191.7 kips Vr,x / Vc,x = 0.264 PASS - Design shear strength exceeds required shear strength Design of members for flexure - Chapter F Required flexural strength;Mr,x = 134.8 kips_ft Yielding - Section F2.1 Nominal flexural strength for yielding - eq F2-1;Mn,yld,x = Mp,x = Fy  Zx = 420.8 kips_ft Lateral-torsional buckling - Section F2.2 Unbraced length;Lb = 6 ft Project Job Ref. Section Existing Column-Beam Frame Sheet no./rev. 16 Calc. by N Date 3/27/2020 Chk'd by Date App'd by Date Limiting unbraced length for yielding - eq F2-5;Lp = 1.76  ry  (E / Fy) = 5.828 ft Distance between flange centroids ;ho = 17.4 in c = 1 rts = 1.98 in Limiting unbraced length for inelastic LTB - eq F2-6;Lr = 1.95  rts  E / (0.7  Fy)  ((J  c / (Sx  ho)) + ((J  c / (Sx  ho))2 + 6.76  (0.7  Fy / E)2)) = 16.946 ft Moment at quarter point of segment;MA = 2.7 kips_ft Moment at center-line of segment;MB = 8.6 kips_ft Moment at three quarter point of segment ;MC = 65.3 kips_ft Maximum moment in segment;Mmax = 134.8 kips_ft LTB modification factor - eq F1-1;Cb = 12.5  Mmax / (2.5  Mmax + 3  MA + 4  MB + 3  MC) = 2.927 Nominal flexural strength for lateral-torsional buckling - eq F2-2 Mn,ltb,x = min(Cb  (Mp,x - (Mp,x - 0.7  Fy  Sx)  (Lb - Lp) / (Lr - Lp)), Mp,x) = 420.8 kips_ft Design flexural strength - F1 Nominal flexural strength;Mn,x = min(Mn,yld,x, Mn,ltb,x) = 420.8 kips_ft Design flexural strength;Mc,x = b  Mn,x = 378.7 kips_ft Mr,x / Mc,x = 0.356 PASS - Design flexural strength exceeds required flexural strength Design of members for combined forces - Chapter H Elastic critical buckling stress;Pey = 2  E  Iy / Lb2 = 2214 kips Nominal flexural strength for combined check;Mn,c,x = min(Mn,yld,x, (1 + Pr / Pey)  Mn,ltb,x) = 420.8 kips_ft Design flexural strength for combined check;Mc,c,x = b  Mn,c,x = 378.7 kips_ft Combined flexure and axial force - eq H1-1b;Pr / (2  Pc) + Mr,x / Mc,c,x = 0.361 PASS - Combined flexure and axial force is within acceptable limits Consider Combination 2 - 1.0D + 1.0L (Service) Check design at end of span Design of members for x-x axis deflection Maximum deflection;x = 0.035 in Allowable deflection;x,Allowable = Lm4_s1 / 360 = 0.2 in x / x,Allowable = 0.176 PASS - Allowable deflection exceeds design deflection Member4 - Span 2 design Section details Section type;W 18x50 (AISC 15th Edn (v15.0)) ASTM steel designation;A992 Steel yield stress;Fy = 50 ksi Steel tensile stress;Fu = 65 ksi Modulus of elasticity;E = 29000 ksi Project Job Ref. Section Existing Column-Beam Frame Sheet no./rev. 17 Calc. by N Date 3/27/2020 Chk'd by Date App'd by Date 7.5" 0.36" 18 " 0 . 5 7 " 0. 5 7 " W 18x50 (AISC 15th Edn (v15.0)) Section depth, d, 18 in Section breadth, bf, 7.5 in Weight of section, Weight, 50 lbf/ft Flange thickness, tf, 0.57 in Web thickness, tw, 0.355 in Area of section, A, 14.7 in2 Radius of gyration about x-axis, rx, 7.38 in Radius of gyration about y-axis, ry, 1.65 in Elastic section modulus about x-axis, Sx, 88.9 in3 Elastic section modulus about y-axis, Sy, 10.7 in3 Plastic section modulus about x-axis, Zx, 101 in3 Plastic section modulus about y-axis, Zy, 16.6 in3 Second moment of area about x-axis, I x, 800 in4 Second moment of area about y-axis, I y, 40.1 in4 Lateral restraint Both flanges have lateral restraint at supports only Classification of sections for local buckling - Section B4 Classification of flanges in flexure - Table B4.1b (case 10) Width to thickness ratio;bf / (2  tf) = 6.58 Limiting ratio for compact section;pff = 0.38  [E / Fy] = 9.15 Limiting ratio for non-compact section;rff = 1.0  [E / Fy] = 24.08;Compact Classification of web in flexure - Table B4.1b (case 15) Width to thickness ratio;(d - 2  k) / tw = 45.23 Limiting ratio for compact section;pwf = 3.76  [E / Fy] = 90.55 Limiting ratio for non-compact section;rwf = 5.70  [E / Fy] = 137.27;Compact Section is compact in flexure Classification of flanges in uniform compression - Table B4.1a (case 1) Width to thickness ratio;bf / (2  tf) = 6.58 Limiting ratio for non-compact section;rfc = 0.56  [E / Fy] = 13.49;Nonslender Classification of web in uniform compression - Table B4.1a (case 5) Width to thickness ratio;(d - 2  k) / tw = 45.23 Limiting ratio for non-compact section;rwc = 1.49  [E / Fy] = 35.88;Slender Section is slender in compression Check design at start of span Design of members for compression - Chapter E Required compressive strength;Pr = 67.2 kips Slenderness limitations and effective length - Section E2 Unbraced length;Lb,x = Lm4_s2 = 20.5 ft Effective length factor;Kx = 1.00 Column slenderness;x = Kx  Lb,x / rx = 33.333 Major axis column slenderness ratio does not exceed recommended limit of 200 Project Job Ref. Section Existing Column-Beam Frame Sheet no./rev. 18 Calc. by N Date 3/27/2020 Chk'd by Date App'd by Date Slenderness limitations and effective length - Section E2 Unbraced length;Lb,y = Lm4_s2_seg1 = 20.5 ft Effective length factor;Ky = 1.00 Column slenderness;y = Ky  Lb,y / ry = 149.091 Minor axis column slenderness ratio does not exceed recommended limit of 200 Flexural buckling of members with slender elements - Section E7 Elastic critical buckling stress - eq E3-4;Fe = 2  E / (Kx  Lb,x / rx)2 = 257.6 ksi Flexural buckling stress;f = 0.658Fy / Fe  Fy = 46.1 ksi Width of unstiffened compression element;h = d - 2  k = 16.056 in Reduced effective height - eq E7-17 ;he = min(1.92  tw  (E / f)  [1 - (0.34 / (h / tw))  (E / f)], h) = 13.9 in Effective area of cross-section;Ae = A - (h - he)  tw = 13.925 in2 Slender stiffened element factor - eq E7-16;Qa = Ae / A = 0.947 Net reduction factor;Q = Qa = 0.947 Elastic critical buckling stress - eq E3-4;Fe,x = 2  E / (Kx  Lb,x / rx)2 = 257.6 ksi Flexural buckling stress - eq E7-2;Fcr,x = Q  [0.658Q  Fy / Fe,x]  Fy = 43.9 ksi Nominal compressive strength for flexural buckling - eq E7-1; Pn,fb,x = Fcr,x  A = 644.7 kips Elastic critical buckling stress - eq E3-4;Fe = 2  E / (Ky  Lb,y / ry)2 = 12.9 ksi Flexural buckling stress;f = 0.877  Fe = 11.3 ksi Width of unstiffened compression element;h = d - 2  k = 16.056 in Reduced effective height - eq E7-17 ;he = h = 16.1 in Effective area of cross-section;Ae = A = 14.7 in2 Slender stiffened element factor - eq E7-16;Qa = Ae / A = 1.000 Net reduction factor;Q = Qa = 1.000 Elastic critical buckling stress - eq E3-4;Fe,y = 2  E / (Ky  Lb,y / ry)2 = 12.9 ksi Flexural buckling stress - eq E7-3;Fcr,y = 0.877  Fe,y = 11.3 ksi Nominal compressive strength for flexural buckling - eq E7-1; Pn,fb,y = Fcr,y  A = 166 kips Torsional and torsional-flexural buckling of members with slender elements - Section E7 Unbraced length;Lb,z = Lm4_s2_seg1_R = 20.5 ft Effective length factor;Kz = 1.00 Flexural-torsional elastic buckling stress - eq E4-4;Fe = (2  E  Cw / (Kz  Lb,z)2 + G  J) / (Ix + Iy) = 33.6 ksi Flexural buckling stress;f = [0.658Fy / Fe]  Fy = 26.8 ksi Width of unstiffened compression element;h = d - 2  k = 16.056 in Reduced effective height - eq E7-17 ;he = h = 16.1 in Effective area of cross-section;Ae = A = 14.7 in2 Slender stiffened element factor - eq E7-16;Qa = Ae / A = 1.000 Net reduction factor;Q = Qa = 1.000 Flexural-torsional buckling stress - eq E7-2;Fcr = Q  [0.658Q  Fy / Fe]  Fy = 26.8 ksi Nominal compressive strength for torsional and flexural-torsional buckling - eq E4-1 ; Pn,ftb = Fcr  A = 394.1 kips Design compressive strength - E1 Nominal compressive strength;Pn = min(Pn,fb,x, Pn,fb,y, Pn,ftb) = 166 kips Project Job Ref. Section Existing Column-Beam Frame Sheet no./rev. 19 Calc. by N Date 3/27/2020 Chk'd by Date App'd by Date Design compressive strength;Pc = c  Pn = 149.4 kips Pr / Pc = 0.450 PASS - Nominal compressive strength exceeds required compressive strength Design of members for shear - Chapter G Required shear strength;Vr,x = 58 kips Web area;Aw = d  tw = 6.39 in2 Web plate buckling coefficient;kv = 5 (d - 2  k) / tw <= 2.24  (E / Fy) Web shear coefficient - eq G2-2;Cv = 1.000 Nominal shear strength - eq G2-1;Vn,x = 0.6  Fy  Aw  Cv = 191.7 kips Resistance factor;v = 1.00 Design shear strength;Vc,x = v  Vn,x = 191.7 kips Vr,x / Vc,x = 0.303 PASS - Design shear strength exceeds required shear strength Design of members for flexure - Chapter F Required flexural strength;Mr,x = 162 kips_ft Yielding - Section F2.1 Nominal flexural strength for yielding - eq F2-1;Mn,yld,x = Mp,x = Fy  Zx = 420.8 kips_ft Lateral-torsional buckling - Section F2.2 Unbraced length;Lb = 20.5 ft Limiting unbraced length for yielding - eq F2-5;Lp = 1.76  ry  (E / Fy) = 5.828 ft Distance between flange centroids ;ho = 17.4 in c = 1 rts = 1.98 in Limiting unbraced length for inelastic LTB - eq F2-6;Lr = 1.95  rts  E / (0.7  Fy)  ((J  c / (Sx  ho)) + ((J  c / (Sx  ho))2 + 6.76  (0.7  Fy / E)2)) = 16.946 ft Moment at quarter point of segment;MA = 61 kips_ft Moment at center-line of segment;MB = 135.3 kips_ft Moment at three quarter point of segment ;MC = 61 kips_ft Maximum moment in segment;Mmax = 162 kips_ft LTB modification factor - eq F1-1;Cb = 12.5  Mmax / (2.5  Mmax + 3  MA + 4  MB + 3  MC) = 1.543 Critical flexural stress - eq F2-4;Fcr = (Cb  2  E / (Lb / rts)2)  (1 + 0.078  (J  c / (Sx  ho))  (Lb / rts)2) = 40.1 ksi Nominal flexural strength for lateral-torsional buckling - eq F2-3 Mn,ltb,x = min(Fcr  Sx, Mp,x) = 297.2 kips_ft Design flexural strength - F1 Nominal flexural strength;Mn,x = min(Mn,yld,x, Mn,ltb,x) = 297.2 kips_ft Design flexural strength;Mc,x = b  Mn,x = 267.4 kips_ft Mr,x / Mc,x = 0.606 PASS - Design flexural strength exceeds required flexural strength Design of members for combined forces - Chapter H Combined flexure and axial force - eq H1-1a;Pr / Pc + 8 / 9  (Mr,x / Mc,x) = 0.988 PASS - Combined flexure and axial force is within acceptable limits Project Job Ref. Section Existing Column-Beam Frame Sheet no./rev. 20 Calc. by N Date 3/27/2020 Chk'd by Date App'd by Date Check design 10ft 3in along span Design of members for compression - Chapter E Required compressive strength;Pr = 67.2 kips Slenderness limitations and effective length - Section E2 Unbraced length;Lb,x = Lm4_s2 = 20.5 ft Effective length factor;Kx = 1.00 Column slenderness;x = Kx  Lb,x / rx = 33.333 Major axis column slenderness ratio does not exceed recommended limit of 200 Slenderness limitations and effective length - Section E2 Unbraced length;Lb,y = Lm4_s2_seg1 = 20.5 ft Effective length factor;Ky = 1.00 Column slenderness;y = Ky  Lb,y / ry = 149.091 Minor axis column slenderness ratio does not exceed recommended limit of 200 Flexural buckling of members with slender elements - Section E7 Elastic critical buckling stress - eq E3-4;Fe = 2  E / (Kx  Lb,x / rx)2 = 257.6 ksi Flexural buckling stress;f = 0.658Fy / Fe  Fy = 46.1 ksi Width of unstiffened compression element;h = d - 2  k = 16.056 in Reduced effective height - eq E7-17 ;he = min(1.92  tw  (E / f)  [1 - (0.34 / (h / tw))  (E / f)], h) = 13.9 in Effective area of cross-section;Ae = A - (h - he)  tw = 13.925 in2 Slender stiffened element factor - eq E7-16;Qa = Ae / A = 0.947 Net reduction factor;Q = Qa = 0.947 Elastic critical buckling stress - eq E3-4;Fe,x = 2  E / (Kx  Lb,x / rx)2 = 257.6 ksi Flexural buckling stress - eq E7-2;Fcr,x = Q  [0.658Q  Fy / Fe,x]  Fy = 43.9 ksi Nominal compressive strength for flexural buckling - eq E7-1; Pn,fb,x = Fcr,x  A = 644.7 kips Elastic critical buckling stress - eq E3-4;Fe = 2  E / (Ky  Lb,y / ry)2 = 12.9 ksi Flexural buckling stress;f = 0.877  Fe = 11.3 ksi Width of unstiffened compression element;h = d - 2  k = 16.056 in Reduced effective height - eq E7-17 ;he = h = 16.1 in Effective area of cross-section;Ae = A = 14.7 in2 Slender stiffened element factor - eq E7-16;Qa = Ae / A = 1.000 Net reduction factor;Q = Qa = 1.000 Elastic critical buckling stress - eq E3-4;Fe,y = 2  E / (Ky  Lb,y / ry)2 = 12.9 ksi Flexural buckling stress - eq E7-3;Fcr,y = 0.877  Fe,y = 11.3 ksi Nominal compressive strength for flexural buckling - eq E7-1; Pn,fb,y = Fcr,y  A = 166 kips Torsional and torsional-flexural buckling of members with slender elements - Section E7 Unbraced length;Lb,z = Lm4_s2_seg1_R = 20.5 ft Effective length factor;Kz = 1.00 Flexural-torsional elastic buckling stress - eq E4-4;Fe = (2  E  Cw / (Kz  Lb,z)2 + G  J) / (Ix + Iy) = 33.6 ksi Flexural buckling stress;f = [0.658Fy / Fe]  Fy = 26.8 ksi Width of unstiffened compression element;h = d - 2  k = 16.056 in Project Job Ref. Section Existing Column-Beam Frame Sheet no./rev. 21 Calc. by N Date 3/27/2020 Chk'd by Date App'd by Date Reduced effective height - eq E7-17 ;he = h = 16.1 in Effective area of cross-section;Ae = A = 14.7 in2 Slender stiffened element factor - eq E7-16;Qa = Ae / A = 1.000 Net reduction factor;Q = Qa = 1.000 Flexural-torsional buckling stress - eq E7-2;Fcr = Q  [0.658Q  Fy / Fe]  Fy = 26.8 ksi Nominal compressive strength for torsional and flexural-torsional buckling - eq E4-1 ; Pn,ftb = Fcr  A = 394.1 kips Design compressive strength - E1 Nominal compressive strength;Pn = min(Pn,fb,x, Pn,fb,y, Pn,ftb) = 166 kips Design compressive strength;Pc = c  Pn = 149.4 kips Pr / Pc = 0.450 PASS - Nominal compressive strength exceeds required compressive strength Design of members for flexure - Chapter F Required flexural strength;Mr,x = 135.3 kips_ft Yielding - Section F2.1 Nominal flexural strength for yielding - eq F2-1;Mn,yld,x = Mp,x = Fy  Zx = 420.8 kips_ft Lateral-torsional buckling - Section F2.2 Unbraced length;Lb = 20.5 ft Limiting unbraced length for yielding - eq F2-5;Lp = 1.76  ry  (E / Fy) = 5.828 ft Distance between flange centroids ;ho = 17.4 in c = 1 rts = 1.98 in Limiting unbraced length for inelastic LTB - eq F2-6;Lr = 1.95  rts  E / (0.7  Fy)  ((J  c / (Sx  ho)) + ((J  c / (Sx  ho))2 + 6.76  (0.7  Fy / E)2)) = 16.946 ft Moment at quarter point of segment;MA = 61 kips_ft Moment at center-line of segment;MB = 135.3 kips_ft Moment at three quarter point of segment ;MC = 61 kips_ft Maximum moment in segment;Mmax = 162 kips_ft LTB modification factor - eq F1-1;Cb = 12.5  Mmax / (2.5  Mmax + 3  MA + 4  MB + 3  MC) = 1.543 Critical flexural stress - eq F2-4;Fcr = (Cb  2  E / (Lb / rts)2)  (1 + 0.078  (J  c / (Sx  ho))  (Lb / rts)2) = 40.1 ksi Nominal flexural strength for lateral-torsional buckling - eq F2-3 Mn,ltb,x = min(Fcr  Sx, Mp,x) = 297.2 kips_ft Design flexural strength - F1 Nominal flexural strength;Mn,x = min(Mn,yld,x, Mn,ltb,x) = 297.2 kips_ft Design flexural strength;Mc,x = b  Mn,x = 267.4 kips_ft Mr,x / Mc,x = 0.506 PASS - Design flexural strength exceeds required flexural strength Design of members for combined forces - Chapter H Combined flexure and axial force - eq H1-1a;Pr / Pc + 8 / 9  (Mr,x / Mc,x) = 0.899 PASS - Combined flexure and axial force is within acceptable limits Project Job Ref. Section Existing Column-Beam Frame Sheet no./rev. 22 Calc. by N Date 3/27/2020 Chk'd by Date App'd by Date Check design at end of span Design of members for compression - Chapter E Required compressive strength;Pr = 67.2 kips Slenderness limitations and effective length - Section E2 Unbraced length;Lb,x = Lm4_s2 = 20.5 ft Effective length factor;Kx = 1.00 Column slenderness;x = Kx  Lb,x / rx = 33.333 Major axis column slenderness ratio does not exceed recommended limit of 200 Slenderness limitations and effective length - Section E2 Unbraced length;Lb,y = Lm4_s2_seg1 = 20.5 ft Effective length factor;Ky = 1.00 Column slenderness;y = Ky  Lb,y / ry = 149.091 Minor axis column slenderness ratio does not exceed recommended limit of 200 Flexural buckling of members with slender elements - Section E7 Elastic critical buckling stress - eq E3-4;Fe = 2  E / (Kx  Lb,x / rx)2 = 257.6 ksi Flexural buckling stress;f = 0.658Fy / Fe  Fy = 46.1 ksi Width of unstiffened compression element;h = d - 2  k = 16.056 in Reduced effective height - eq E7-17 ;he = min(1.92  tw  (E / f)  [1 - (0.34 / (h / tw))  (E / f)], h) = 13.9 in Effective area of cross-section;Ae = A - (h - he)  tw = 13.925 in2 Slender stiffened element factor - eq E7-16;Qa = Ae / A = 0.947 Net reduction factor;Q = Qa = 0.947 Elastic critical buckling stress - eq E3-4;Fe,x = 2  E / (Kx  Lb,x / rx)2 = 257.6 ksi Flexural buckling stress - eq E7-2;Fcr,x = Q  [0.658Q  Fy / Fe,x]  Fy = 43.9 ksi Nominal compressive strength for flexural buckling - eq E7-1; Pn,fb,x = Fcr,x  A = 644.7 kips Elastic critical buckling stress - eq E3-4;Fe = 2  E / (Ky  Lb,y / ry)2 = 12.9 ksi Flexural buckling stress;f = 0.877  Fe = 11.3 ksi Width of unstiffened compression element;h = d - 2  k = 16.056 in Reduced effective height - eq E7-17 ;he = h = 16.1 in Effective area of cross-section;Ae = A = 14.7 in2 Slender stiffened element factor - eq E7-16;Qa = Ae / A = 1.000 Net reduction factor;Q = Qa = 1.000 Elastic critical buckling stress - eq E3-4;Fe,y = 2  E / (Ky  Lb,y / ry)2 = 12.9 ksi Flexural buckling stress - eq E7-3;Fcr,y = 0.877  Fe,y = 11.3 ksi Nominal compressive strength for flexural buckling - eq E7-1; Pn,fb,y = Fcr,y  A = 166 kips Torsional and torsional-flexural buckling of members with slender elements - Section E7 Unbraced length;Lb,z = Lm4_s2_seg1_R = 20.5 ft Effective length factor;Kz = 1.00 Flexural-torsional elastic buckling stress - eq E4-4;Fe = (2  E  Cw / (Kz  Lb,z)2 + G  J) / (Ix + Iy) = 33.6 ksi Flexural buckling stress;f = [0.658Fy / Fe]  Fy = 26.8 ksi Width of unstiffened compression element;h = d - 2  k = 16.056 in Project Job Ref. Section Existing Column-Beam Frame Sheet no./rev. 23 Calc. by N Date 3/27/2020 Chk'd by Date App'd by Date Reduced effective height - eq E7-17 ;he = h = 16.1 in Effective area of cross-section;Ae = A = 14.7 in2 Slender stiffened element factor - eq E7-16;Qa = Ae / A = 1.000 Net reduction factor;Q = Qa = 1.000 Flexural-torsional buckling stress - eq E7-2;Fcr = Q  [0.658Q  Fy / Fe]  Fy = 26.8 ksi Nominal compressive strength for torsional and flexural-torsional buckling - eq E4-1 ; Pn,ftb = Fcr  A = 394.1 kips Design compressive strength - E1 Nominal compressive strength;Pn = min(Pn,fb,x, Pn,fb,y, Pn,ftb) = 166 kips Design compressive strength;Pc = c  Pn = 149.4 kips Pr / Pc = 0.450 PASS - Nominal compressive strength exceeds required compressive strength Design of members for shear - Chapter G Required shear strength;Vr,x = 58 kips Web area;Aw = d  tw = 6.39 in2 Web plate buckling coefficient;kv = 5 (d - 2  k) / tw <= 2.24  (E / Fy) Web shear coefficient - eq G2-2;Cv = 1.000 Nominal shear strength - eq G2-1;Vn,x = 0.6  Fy  Aw  Cv = 191.7 kips Resistance factor;v = 1.00 Design shear strength;Vc,x = v  Vn,x = 191.7 kips Vr,x / Vc,x = 0.303 PASS - Design shear strength exceeds required shear strength Design of members for flexure - Chapter F Required flexural strength;Mr,x = 162 kips_ft Yielding - Section F2.1 Nominal flexural strength for yielding - eq F2-1;Mn,yld,x = Mp,x = Fy  Zx = 420.8 kips_ft Lateral-torsional buckling - Section F2.2 Unbraced length;Lb = 20.5 ft Limiting unbraced length for yielding - eq F2-5;Lp = 1.76  ry  (E / Fy) = 5.828 ft Distance between flange centroids ;ho = 17.4 in c = 1 rts = 1.98 in Limiting unbraced length for inelastic LTB - eq F2-6;Lr = 1.95  rts  E / (0.7  Fy)  ((J  c / (Sx  ho)) + ((J  c / (Sx  ho))2 + 6.76  (0.7  Fy / E)2)) = 16.946 ft Moment at quarter point of segment;MA = 61 kips_ft Moment at center-line of segment;MB = 135.3 kips_ft Moment at three quarter point of segment ;MC = 61 kips_ft Maximum moment in segment;Mmax = 162 kips_ft LTB modification factor - eq F1-1;Cb = 12.5  Mmax / (2.5  Mmax + 3  MA + 4  MB + 3  MC) = 1.543 Critical flexural stress - eq F2-4;Fcr = (Cb  2  E / (Lb / rts)2)  (1 + 0.078  (J  c / (Sx  ho))  (Lb / rts)2) = 40.1 ksi Nominal flexural strength for lateral-torsional buckling - eq F2-3 Project Job Ref. Section Existing Column-Beam Frame Sheet no./rev. 24 Calc. by N Date 3/27/2020 Chk'd by Date App'd by Date Mn,ltb,x = min(Fcr  Sx, Mp,x) = 297.2 kips_ft Design flexural strength - F1 Nominal flexural strength;Mn,x = min(Mn,yld,x, Mn,ltb,x) = 297.2 kips_ft Design flexural strength;Mc,x = b  Mn,x = 267.4 kips_ft Mr,x / Mc,x = 0.606 PASS - Design flexural strength exceeds required flexural strength Design of members for combined forces - Chapter H Combined flexure and axial force - eq H1-1a;Pr / Pc + 8 / 9  (Mr,x / Mc,x) = 0.988 PASS - Combined flexure and axial force is within acceptable limits Consider Combination 2 - 1.0D + 1.0L (Service) Check design 10ft 3in along span Design of members for x-x axis deflection Maximum deflection;x = 0.303 in Allowable deflection;x,Allowable = Lm4_s2 / 360 = 0.683 in x / x,Allowable = 0.444 PASS - Allowable deflection exceeds design deflection Member4 - Span 3 design Section details Section type;W 18x50 (AISC 15th Edn (v15.0)) ASTM steel designation;A992 Steel yield stress;Fy = 50 ksi Steel tensile stress;Fu = 65 ksi Modulus of elasticity;E = 29000 ksi 7.5" 0.36" 18 " 0 . 5 7 " 0. 5 7 " W 18x50 (AISC 15th Edn (v15.0)) Section depth, d, 18 in Section breadth, bf, 7.5 in Weight of section, Weight, 50 lbf/ft Flange thickness, tf, 0.57 in Web thickness, tw, 0.355 in Area of section, A, 14.7 in2 Radius of gyration about x-axis, rx, 7.38 in Radius of gyration about y-axis, ry, 1.65 in Elastic section modulus about x-axis, Sx, 88.9 in3 Elastic section modulus about y-axis, Sy, 10.7 in3 Plastic section modulus about x-axis, Zx, 101 in3 Plastic section modulus about y-axis, Zy, 16.6 in3 Second moment of area about x-axis, I x, 800 in4 Second moment of area about y-axis, I y, 40.1 in4 Lateral restraint Both flanges have lateral restraint at supports only Classification of sections for local buckling - Section B4 Project Job Ref. Section Existing Column-Beam Frame Sheet no./rev. 25 Calc. by N Date 3/27/2020 Chk'd by Date App'd by Date Classification of flanges in flexure - Table B4.1b (case 10) Width to thickness ratio;bf / (2  tf) = 6.58 Limiting ratio for compact section;pff = 0.38  [E / Fy] = 9.15 Limiting ratio for non-compact section;rff = 1.0  [E / Fy] = 24.08;Compact Classification of web in flexure - Table B4.1b (case 15) Width to thickness ratio;(d - 2  k) / tw = 45.23 Limiting ratio for compact section;pwf = 3.76  [E / Fy] = 90.55 Limiting ratio for non-compact section;rwf = 5.70  [E / Fy] = 137.27;Compact Section is compact in flexure Classification of flanges in uniform compression - Table B4.1a (case 1) Width to thickness ratio;bf / (2  tf) = 6.58 Limiting ratio for non-compact section;rfc = 0.56  [E / Fy] = 13.49;Nonslender Classification of web in uniform compression - Table B4.1a (case 5) Width to thickness ratio;(d - 2  k) / tw = 45.23 Limiting ratio for non-compact section;rwc = 1.49  [E / Fy] = 35.88;Slender Section is slender in compression Check design at start of span Design of members for tension - Chapter D Required tensile strength;Pr = 0.751 kips Slenderness limitations - Section D1 Slenderness ratio; = max(Lb,x / rx, Lb,y / ry) = 149.091 Tension member slenderness ratio does not exceed recommended limit of 300 Design tensile strength - D2 Nominal tensile yielding strength;Pn,yld = Fy  A = 735 kips Nominal tensile rupture strength;Pn,r = Fu  Ae = 955.5 kips Design tensile strength;Pc = min(t,y  Pn,yld, t,r  Pn,r) = 661.5 kips Pr / Pc = 0.001 PASS - Nominal tensile strength exceeds required tensile strength Design of members for shear - Chapter G Required shear strength;Vr,x = 43.6 kips Web area;Aw = d  tw = 6.39 in2 Web plate buckling coefficient;kv = 5 (d - 2  k) / tw <= 2.24  (E / Fy) Web shear coefficient - eq G2-2;Cv = 1.000 Nominal shear strength - eq G2-1;Vn,x = 0.6  Fy  Aw  Cv = 191.7 kips Resistance factor;v = 1.00 Design shear strength;Vc,x = v  Vn,x = 191.7 kips Vr,x / Vc,x = 0.227 PASS - Design shear strength exceeds required shear strength Design of members for flexure - Chapter F Required flexural strength;Mr,x = 130.1 kips_ft Project Job Ref. Section Existing Column-Beam Frame Sheet no./rev. 26 Calc. by N Date 3/27/2020 Chk'd by Date App'd by Date Yielding - Section F2.1 Nominal flexural strength for yielding - eq F2-1;Mn,yld,x = Mp,x = Fy  Zx = 420.8 kips_ft Lateral-torsional buckling - Section F2.2 Unbraced length;Lb = 6 ft Limiting unbraced length for yielding - eq F2-5;Lp = 1.76  ry  (E / Fy) = 5.828 ft Distance between flange centroids ;ho = 17.4 in c = 1 rts = 1.98 in Limiting unbraced length for inelastic LTB - eq F2-6;Lr = 1.95  rts  E / (0.7  Fy)  ((J  c / (Sx  ho)) + ((J  c / (Sx  ho))2 + 6.76  (0.7  Fy / E)2)) = 16.946 ft Moment at quarter point of segment;MA = 71.1 kips_ft Moment at center-line of segment;MB = 24.9 kips_ft Moment at three quarter point of segment ;MC = 9.2 kips_ft Maximum moment in segment;Mmax = 130.1 kips_ft LTB modification factor - eq F1-1;Cb = 12.5  Mmax / (2.5  Mmax + 3  MA + 4  MB + 3  MC) = 2.443 Nominal flexural strength for lateral-torsional buckling - eq F2-2 Mn,ltb,x = min(Cb  (Mp,x - (Mp,x - 0.7  Fy  Sx)  (Lb - Lp) / (Lr - Lp)), Mp,x) = 420.8 kips_ft Design flexural strength - F1 Nominal flexural strength;Mn,x = min(Mn,yld,x, Mn,ltb,x) = 420.8 kips_ft Design flexural strength;Mc,x = b  Mn,x = 378.7 kips_ft Mr,x / Mc,x = 0.343 PASS - Design flexural strength exceeds required flexural strength Design of members for combined forces - Chapter H Elastic critical buckling stress;Pey = 2  E  Iy / Lb2 = 2214 kips Nominal flexural strength for combined check;Mn,c,x = min(Mn,yld,x, (1 + Pr / Pey)  Mn,ltb,x) = 420.8 kips_ft Design flexural strength for combined check;Mc,c,x = b  Mn,c,x = 378.7 kips_ft Combined flexure and axial force - eq H1-1b;Pr / (2  Pc) + Mr,x / Mc,c,x = 0.344 PASS - Combined flexure and axial force is within acceptable limits Consider Combination 2 - 1.0D + 1.0L (Service) Check design at start of span Design of members for x-x axis deflection Maximum deflection;x = 0.035 in Allowable deflection;x,Allowable = Lm4_s3 / 360 = 0.2 in x / x,Allowable = 0.176 PASS - Allowable deflection exceeds design deflection Member5 design Section details Section type;HSS 12x8x1/4 (AISC 15th Edn (v15.0)) ASTM steel designation;User defined Steel yield stress;Fy = 42 ksi Project Job Ref. Section Existing Column-Beam Frame Sheet no./rev. 27 Calc. by N Date 3/27/2020 Chk'd by Date App'd by Date Steel tensile stress;Fu = 65 ksi Modulus of elasticity;E = 29000 ksi 8" 0.23" 12 " HSS 12x8x1/4 (AISC 15th Edn (v15.0)) Section depth, d, 12 in Section breadth, bf, 8 in Weight of section, Weight, 32.6 lbf/ft Section thickness, t, 0.233 in Area of section, A, 9 in2 Radius of gyration about x-axis, r x, 4.53 in Radius of gyration about y-axis, r y, 3.32 in Elastic section modulus about x-axis, Sx, 30.6 in3 Elastic section modulus about y-axis, Sy, 24.7 in3 Plastic section modulus about x-axis, Zx, 36.6 in3 Plastic section modulus about y-axis, Zy, 27.8 in3 Second moment of area about x-axis, I x, 184 in4 Second moment of area about y-axis, I y, 98.8 in4 Lateral restraint Both flanges have lateral restraint at supports only Classification of sections for local buckling - Section B4 Classification of flanges in flexure - Table B4.1b (case 17) Width to thickness ratio;(bf - 3  t) / t = 31.33 Limiting ratio for compact section;pff = 1.12  [E / Fy] = 29.43 Limiting ratio for non-compact section;rff = 1.40  [E / Fy] = 36.79;Noncompact Classification of web in flexure - Table B4.1b (case 19) Width to thickness ratio;(d - 3  t) / t = 48.50 Limiting ratio for compact section;pwf = 2.42  [E / Fy] = 63.59 Limiting ratio for non-compact section;rwf = 5.70  [E / Fy] = 149.78;Compact Section is noncompact in flexure Classification of flanges in uniform compression - Table B4.1a (case 6) Width to thickness ratio;(bf - 3  t) / t = 31.33 Limiting ratio for non-compact section;rfc = 1.40  [E / Fy] = 36.79;Nonslender Classification of web in uniform compression - Table B4.1a (case 6) Width to thickness ratio;(d - 3  t) / t = 48.50 Limiting ratio for non-compact section;rwc = 1.40  [E / Fy] = 36.79;Slender Section is slender in compression Check design at start of span Design of members for compression - Chapter E Required compressive strength;Pr = "Expresso error (169)" kips Check design at end of span Design of members for tension - Chapter D Required tensile strength;Pr = 0.082 kips Project Job Ref. Section Existing Column-Beam Frame Sheet no./rev. 28 Calc. by N Date 3/27/2020 Chk'd by Date App'd by Date Slenderness limitations - Section D1 Slenderness ratio; = max(Lb,x / rx, Lb,y / ry) = ?<Error: invalid expression> Tension member slenderness ratio does not exceed recommended limit of 300 Design tensile strength - D2 Nominal tensile yielding strength;Pn,yld = Fy  A = 376.32 kips Nominal tensile rupture strength;Pn,r = Fu  Ae = 955.5 kips Design tensile strength;Pc = min(t,y  Pn,yld, t,r  Pn,r) = 338.688 kips Pr / Pc = 0 PASS - Nominal tensile strength exceeds required tensile strength Design of members for shear - Chapter G Required shear strength;Vr,x = 0.9 kips Web area;Aw = 2  (d - 3  t)  t = 5.266 in2 Web plate buckling coefficient;kv = 5 (d - 3  t) / t <= 1.10  (kv  E / Fy) Web shear coefficient - eq G2-3;Cv = 1.000 Nominal shear strength - eq G2-1;Vn,x = 0.6  Fy  Aw  Cv = 132.7 kips Resistance factor;v = 0.90 Design shear strength;Vc,x = v  Vn,x = 119.4 kips Vr,x / Vc,x = 0.008 PASS - Design shear strength exceeds required shear strength Design of members for flexure - Chapter F Required flexural strength;Mr,x = 4 kips_ft Yielding - Section F7.1 Nominal flexural strength for yielding - eq F7-1;Mn,yld,x = Mp,x = Fy  Zx = 128.1 kips_ft Compression flange local buckling - Section F7.2 Nominal flexural strength for compression flange local buckling - eq F7-2 ; Mn,flb,x = min(Mp,x - (Mp,x - Fy  Sx)  (3.57  (bf - 3  t) / t  (Fy / E) - 4.0), Mp,x) = 122.7 kips_ft Web local buckling - Section F7.3 Nominal flexural strength for web local buckling - eq F7-5; Mn,wlb,x = Mp,x = 128.1 kips_ft Design flexural strength - F1 Nominal flexural strength;Mn,x = min(Mn,yld,x, Mn,flb,x) = 122.7 kips_ft Design flexural strength;Mc,x = b  Mn,x = 110.4 kips_ft Mr,x / Mc,x = 0.036 PASS - Design flexural strength exceeds required flexural strength Design of members for combined forces - Chapter H Combined flexure and axial force - eq H1-1b;Pr / (2  Pc) + Mr,x / Mc,x = 0.036 PASS - Combined flexure and axial force is within acceptable limits Project Job Ref. Section Existing Column-Beam Frame Sheet no./rev. 29 Calc. by N Date 3/27/2020 Chk'd by Date App'd by Date Consider Combination 10 - 1.0D + 0.5L + 0.5S + 0.7W (Service) Check design 3ft 10.859in along span Design of members for x-x axis deflection Maximum deflection;x = 0.001 in Allowable deflection;x,Allowable = Lm5_s1 / 360 = 0.2 in x / x,Allowable = 0.007 PASS - Allowable deflection exceeds design deflection Member6 design Section details Section type;HSS 12x8x1/4 (AISC 15th Edn (v15.0)) ASTM steel designation;User defined Steel yield stress;Fy = 42 ksi Steel tensile stress;Fu = 65 ksi Modulus of elasticity;E = 29000 ksi 8" 0.23" 1 2 " HSS 12x8x1/4 (AISC 15th Edn (v15.0)) Section depth, d, 12 in Section breadth, bf, 8 in Weight of section, Weight, 32.6 lbf/ft Section thickness, t, 0.233 in Area of section, A, 9 in2 Radius of gyration about x-axis, r x, 4.53 in Radius of gyration about y-axis, r y, 3.32 in Elastic section modulus about x-axis, Sx, 30.6 in3 Elastic section modulus about y-axis, Sy, 24.7 in3 Plastic section modulus about x-axis, Zx, 36.6 in3 Plastic section modulus about y-axis, Zy, 27.8 in3 Second moment of area about x-axis, I x, 184 in4 Second moment of area about y-axis, I y, 98.8 in4 Lateral restraint Both flanges have lateral restraint at supports only Classification of sections for local buckling - Section B4 Classification of flanges in flexure - Table B4.1b (case 17) Width to thickness ratio;(bf - 3  t) / t = 31.33 Limiting ratio for compact section;pff = 1.12  [E / Fy] = 29.43 Limiting ratio for non-compact section;rff = 1.40  [E / Fy] = 36.79;Noncompact Classification of web in flexure - Table B4.1b (case 19) Width to thickness ratio;(d - 3  t) / t = 48.50 Limiting ratio for compact section;pwf = 2.42  [E / Fy] = 63.59 Limiting ratio for non-compact section;rwf = 5.70  [E / Fy] = 149.78;Compact Section is noncompact in flexure Project Job Ref. Section Existing Column-Beam Frame Sheet no./rev. 30 Calc. by N Date 3/27/2020 Chk'd by Date App'd by Date Classification of flanges in uniform compression - Table B4.1a (case 6) Width to thickness ratio;(bf - 3  t) / t = 31.33 Limiting ratio for non-compact section;rfc = 1.40  [E / Fy] = 36.79;Nonslender Classification of web in uniform compression - Table B4.1a (case 6) Width to thickness ratio;(d - 3  t) / t = 48.50 Limiting ratio for non-compact section;rwc = 1.40  [E / Fy] = 36.79;Slender Section is slender in compression Check design at start of span Design of members for compression - Chapter E Required compressive strength;Pr = "Expresso error (169)" kips Check design at end of span Design of members for compression - Chapter E Required compressive strength;Pr = 129.7 kips Slenderness limitations and effective length - Section E2 Unbraced length;Lb,x = Lm6_s1 = 5.831 ft Effective length factor;Kx = 1.00 Column slenderness;x = Kx  Lb,x / rx = 15.446 Major axis column slenderness ratio does not exceed recommended limit of 200 Slenderness limitations and effective length - Section E2 Unbraced length;Lb,y = Lm6_s1_seg1 = 5.831 ft Effective length factor;Ky = 1.00 Column slenderness;y = Ky  Lb,y / ry = 21.076 Minor axis column slenderness ratio does not exceed recommended limit of 200 Flexural buckling of members with slender elements - Section E7 Width of stiffened element b = bf - 3 t = 7.301 in Reduced effective width - eq E7-18 be = b = 7.3 in Height of stiffened compression element h = d - 3 t = 11.301 in Reduced effective height - eq E7-18 he = min(1.92  t  (E / f)  [1 - (0.38 / (h / t))  (E / f)], h) = 10.9 in Effective area of cross-section Ae = A - 2  (h - he)  t = 8.783 in2 Slender stiffened element factor - eq E7-16 Qa = Ae / A = 0.980 Net reduction factor;Q = Qa = 0.980 Elastic critical buckling stress - eq E3-4;Fe,x = 2  E / (Kx  Lb,x / rx)2 = 1199.6 ksi Flexural buckling stress - eq E7-2;Fcr,x = Q  [0.658Q  Fy / Fe,x]  Fy = 40.6 ksi Nominal compressive strength for flexural buckling - eq E7-1; Pn,fb,x = Fcr,x  A = 363.6 kips Width of stiffened element b = bf - 3 t = 7.301 in Reduced effective width - eq E7-18 be = b = 7.3 in Height of stiffened compression element h = d - 3 t = 11.301 in Reduced effective height - eq E7-18 he = min(1.92  t  (E / f)  [1 - (0.38 / (h / t))  (E / f)], h) = 10.9 in Effective area of cross-section Ae = A - 2  (h - he)  t = 8.783 in2 Slender stiffened element factor - eq E7-16 Qa = Ae / A = 0.980 Project Job Ref. Section Existing Column-Beam Frame Sheet no./rev. 31 Calc. by N Date 3/27/2020 Chk'd by Date App'd by Date Net reduction factor;Q = Qa = 0.980 Elastic critical buckling stress - eq E3-4;Fe,y = 2  E / (Ky  Lb,y / ry)2 = 644.4 ksi Flexural buckling stress - eq E7-2;Fcr,y = Q  [0.658Q  Fy / Fe,y]  Fy = 40.1 ksi Nominal compressive strength for flexural buckling - eq E7-1; Pn,fb,y = Fcr,y  A = 359.2 kips Design compressive strength - E1 Nominal compressive strength;Pn = min(Pn,fb,x, Pn,fb,y) = 359.2 kips Design compressive strength;Pc = c  Pn = 323.3 kips Pr / Pc = 0.401 PASS - Nominal compressive strength exceeds required compressive strength Design of members for shear - Chapter G Required shear strength;Vr,x = 5.2 kips Web area;Aw = 2  (d - 3  t)  t = 5.266 in2 Web plate buckling coefficient;kv = 5 (d - 3  t) / t <= 1.10  (kv  E / Fy) Web shear coefficient - eq G2-3;Cv = 1.000 Nominal shear strength - eq G2-1;Vn,x = 0.6  Fy  Aw  Cv = 132.7 kips Resistance factor;v = 0.90 Design shear strength;Vc,x = v  Vn,x = 119.4 kips Vr,x / Vc,x = 0.043 PASS - Design shear strength exceeds required shear strength Design of members for flexure - Chapter F Required flexural strength;Mr,x = 28.9 kips_ft Yielding - Section F7.1 Nominal flexural strength for yielding - eq F7-1;Mn,yld,x = Mp,x = Fy  Zx = 128.1 kips_ft Compression flange local buckling - Section F7.2 Nominal flexural strength for compression flange local buckling - eq F7-2 ; Mn,flb,x = min(Mp,x - (Mp,x - Fy  Sx)  (3.57  (bf - 3  t) / t  (Fy / E) - 4.0), Mp,x) = 122.7 kips_ft Web local buckling - Section F7.3 Nominal flexural strength for web local buckling - eq F7-5; Mn,wlb,x = Mp,x = 128.1 kips_ft Design flexural strength - F1 Nominal flexural strength;Mn,x = min(Mn,yld,x, Mn,flb,x) = 122.7 kips_ft Design flexural strength;Mc,x = b  Mn,x = 110.4 kips_ft Mr,x / Mc,x = 0.262 PASS - Design flexural strength exceeds required flexural strength Design of members for combined forces - Chapter H Combined flexure and axial force - eq H1-1a;Pr / Pc + 8 / 9  (Mr,x / Mc,x) = 0.634 PASS - Combined flexure and axial force is within acceptable limits Project Job Ref. Section Existing Column-Beam Frame Sheet no./rev. 32 Calc. by N Date 3/27/2020 Chk'd by Date App'd by Date Consider Combination 2 - 1.0D + 1.0L (Service) Check design at end of span Design of members for x-x axis deflection Maximum deflection;x = 0.03 in Allowable deflection;x,Allowable = Lm6_s1 / 360 = 0.194 in x / x,Allowable = 0.153 PASS - Allowable deflection exceeds design deflection Member7 design Section details Section type;HSS 12x8x1/4 (AISC 15th Edn (v15.0)) ASTM steel designation;User defined Steel yield stress;Fy = 42 ksi Steel tensile stress;Fu = 65 ksi Modulus of elasticity;E = 29000 ksi 8" 0.23" 1 2 " HSS 12x8x1/4 (AISC 15th Edn (v15.0)) Section depth, d, 12 in Section breadth, bf, 8 in Weight of section, Weight, 32.6 lbf/ft Section thickness, t, 0.233 in Area of section, A, 9 in2 Radius of gyration about x-axis, r x, 4.53 in Radius of gyration about y-axis, r y, 3.32 in Elastic section modulus about x-axis, Sx, 30.6 in3 Elastic section modulus about y-axis, Sy, 24.7 in3 Plastic section modulus about x-axis, Zx, 36.6 in3 Plastic section modulus about y-axis, Zy, 27.8 in3 Second moment of area about x-axis, I x, 184 in4 Second moment of area about y-axis, I y, 98.8 in4 Lateral restraint Both flanges have lateral restraint at supports only Consider Combination 5 - 1.2D + 1.0L + 1.6S (Strength) Classification of sections for local buckling - Section B4 Classification of flanges in flexure - Table B4.1b (case 17) Width to thickness ratio;(bf - 3  t) / t = 31.33 Limiting ratio for compact section;pff = 1.12  [E / Fy] = 29.43 Limiting ratio for non-compact section;rff = 1.40  [E / Fy] = 36.79;Noncompact Classification of web in flexure - Table B4.1b (case 19) Width to thickness ratio;(d - 3  t) / t = 48.50 Limiting ratio for compact section;pwf = 2.42  [E / Fy] = 63.59 Limiting ratio for non-compact section;rwf = 5.70  [E / Fy] = 149.78;Compact Project Job Ref. Section Existing Column-Beam Frame Sheet no./rev. 33 Calc. by N Date 3/27/2020 Chk'd by Date App'd by Date Section is noncompact in flexure Classification of flanges in uniform compression - Table B4.1a (case 6) Width to thickness ratio;(bf - 3  t) / t = 31.33 Limiting ratio for non-compact section;rfc = 1.40  [E / Fy] = 36.79;Nonslender Classification of web in uniform compression - Table B4.1a (case 6) Width to thickness ratio;(d - 3  t) / t = 48.50 Limiting ratio for non-compact section;rwc = 1.40  [E / Fy] = 36.79;Slender Section is slender in compression Check design at start of span Design of members for compression - Chapter E Required compressive strength;Pr = "Expresso error (169)" kips Check design at end of span Design of members for compression - Chapter E Required compressive strength;Pr = 26 kips Slenderness limitations and effective length - Section E2 Unbraced length;Lb,x = Lm7_s1 = 5.831 ft Effective length factor;Kx = 1.00 Column slenderness;x = Kx  Lb,x / rx = 15.446 Major axis column slenderness ratio does not exceed recommended limit of 200 Slenderness limitations and effective length - Section E2 Unbraced length;Lb,y = Lm7_s1_seg1 = 5.831 ft Effective length factor;Ky = 1.00 Column slenderness;y = Ky  Lb,y / ry = 21.076 Minor axis column slenderness ratio does not exceed recommended limit of 200 Flexural buckling of members with slender elements - Section E7 Width of stiffened element b = bf - 3 t = 7.301 in Reduced effective width - eq E7-18 be = b = 7.3 in Height of stiffened compression element h = d - 3 t = 11.301 in Reduced effective height - eq E7-18 he = min(1.92  t  (E / f)  [1 - (0.38 / (h / t))  (E / f)], h) = 10.9 in Effective area of cross-section Ae = A - 2  (h - he)  t = 8.783 in2 Slender stiffened element factor - eq E7-16 Qa = Ae / A = 0.980 Net reduction factor;Q = Qa = 0.980 Elastic critical buckling stress - eq E3-4;Fe,x = 2  E / (Kx  Lb,x / rx)2 = 1199.6 ksi Flexural buckling stress - eq E7-2;Fcr,x = Q  [0.658Q  Fy / Fe,x]  Fy = 40.6 ksi Nominal compressive strength for flexural buckling - eq E7-1; Pn,fb,x = Fcr,x  A = 363.6 kips Width of stiffened element b = bf - 3 t = 7.301 in Reduced effective width - eq E7-18 be = b = 7.3 in Height of stiffened compression element h = d - 3 t = 11.301 in Reduced effective height - eq E7-18 he = min(1.92  t  (E / f)  [1 - (0.38 / (h / t))  (E / f)], h) = 10.9 in Effective area of cross-section Ae = A - 2  (h - he)  t = 8.783 in2 Project Job Ref. Section Existing Column-Beam Frame Sheet no./rev. 34 Calc. by N Date 3/27/2020 Chk'd by Date App'd by Date Slender stiffened element factor - eq E7-16 Qa = Ae / A = 0.980 Net reduction factor;Q = Qa = 0.980 Elastic critical buckling stress - eq E3-4;Fe,y = 2  E / (Ky  Lb,y / ry)2 = 644.4 ksi Flexural buckling stress - eq E7-2;Fcr,y = Q  [0.658Q  Fy / Fe,y]  Fy = 40.1 ksi Nominal compressive strength for flexural buckling - eq E7-1; Pn,fb,y = Fcr,y  A = 359.2 kips Design compressive strength - E1 Nominal compressive strength;Pn = min(Pn,fb,x, Pn,fb,y) = 359.2 kips Design compressive strength;Pc = c  Pn = 323.3 kips Pr / Pc = 0.080 PASS - Nominal compressive strength exceeds required compressive strength Design of members for shear - Chapter G Required shear strength;Vr,x = 3.4 kips Web area;Aw = 2  (d - 3  t)  t = 5.266 in2 Web plate buckling coefficient;kv = 5 (d - 3  t) / t <= 1.10  (kv  E / Fy) Web shear coefficient - eq G2-3;Cv = 1.000 Nominal shear strength - eq G2-1;Vn,x = 0.6  Fy  Aw  Cv = 132.7 kips Resistance factor;v = 0.90 Design shear strength;Vc,x = v  Vn,x = 119.4 kips Vr,x / Vc,x = 0.028 PASS - Design shear strength exceeds required shear strength Design of members for flexure - Chapter F Required flexural strength;Mr,x = 16.1 kips_ft Yielding - Section F7.1 Nominal flexural strength for yielding - eq F7-1;Mn,yld,x = Mp,x = Fy  Zx = 128.1 kips_ft Compression flange local buckling - Section F7.2 Nominal flexural strength for compression flange local buckling - eq F7-2 ; Mn,flb,x = min(Mp,x - (Mp,x - Fy  Sx)  (3.57  (bf - 3  t) / t  (Fy / E) - 4.0), Mp,x) = 122.7 kips_ft Web local buckling - Section F7.3 Nominal flexural strength for web local buckling - eq F7-5; Mn,wlb,x = Mp,x = 128.1 kips_ft Design flexural strength - F1 Nominal flexural strength;Mn,x = min(Mn,yld,x, Mn,flb,x) = 122.7 kips_ft Design flexural strength;Mc,x = b  Mn,x = 110.4 kips_ft Mr,x / Mc,x = 0.146 PASS - Design flexural strength exceeds required flexural strength Design of members for combined forces - Chapter H Combined flexure and axial force - eq H1-1b;Pr / (2  Pc) + Mr,x / Mc,x = 0.186 PASS - Combined flexure and axial force is within acceptable limits Project Job Ref. Section Existing Column-Beam Frame Sheet no./rev. 35 Calc. by N Date 3/27/2020 Chk'd by Date App'd by Date Consider Combination 2 - 1.0D + 1.0L (Service) Check design at end of span Design of members for x-x axis deflection Maximum deflection;x = 0.015 in Allowable deflection;x,Allowable = Lm7_s1 / 360 = 0.194 in x / x,Allowable = 0.078 PASS - Allowable deflection exceeds design deflection Project Job Ref. Section New Column/Beam Frame Sheet no./rev. 1 Calc. by N Date 3/27/2020 Chk'd by Date App'd by Date STEEL MEMBER ANALYSIS & DESIGN (AISC 360) In accordance with AISC360 14th Edition published 2010 using the LRFD method Tedds calculation version 4.3.04 ANALYSIS Tedds calculation version 1.0.27 Geometry Geometry (ft) - Steel (AISC) Me m b e r 1 Member2 Me m b e r 3 111 2 26 3 11 1 2 3 4 X Z Materials Name Density Youngs Modulus Shear Modulus Thermal Coefficient (lbm/ft3) ksi ksi C-1 Steel (AISC) 490 29000 11200 0.000012 Sections Name Area Moment of inertia Shear area parallel to Major Minor Minor Major (in2) (in 4) (in 4) (in 2) (in 2) HSS 8x8x3/8 10 100 100 5 5 W 16x77 23 1110 138 8 14 Nodes Node Co-ordinates Freedom Coordinate system Spring X Z X Z Rot. Name Angle X Z Rot. (ft) (ft)()(kips/ft) (kips/ft)kip_ft/ 1 0 0 Fixed Fixed Fixed 0 0 0 0 2 0 11 Free Free Free 0 0 0 0 3 26 11 Free Free Free 0 0 0 0 4 26 0 Fixed Fixed Fixed 0 0 0 0 Elements Project Job Ref. Section New Column/Beam Frame Sheet no./rev. 2 Calc. by N Date 3/27/2020 Chk'd by Date App'd by Date Element Length Nodes Section Material Releases Rotated (ft) Start End Start moment End moment Axial 1 11 1 2 HSS 8x8x3/8 Steel (AISC) Fixed Fixed Fixed 2 26 2 3 W 16x77 Steel (AISC) Fixed Fixed Fixed 3 11 3 4 HSS 8x8x3/8 Steel (AISC) Fixed Fixed Fixed Members Name Elements Start End Member1 1 1 Member2 2 2 Member3 3 3 Loading Self weight included Dead - Loading (kips/ft,kips) 1. 5 1. 5 7 7 Me m b e r 1 Member2 Me m b e r 3 X Z Live - Loading (kips/ft) 2 . 1 2 . 1 Me m b e r 1 Member2 Me m b e r 3 X Z Project Job Ref. Section New Column/Beam Frame Sheet no./rev. 3 Calc. by N Date 3/27/2020 Chk'd by Date App'd by Date Wind - Loading (kips) 14 M e m b e r 1 Member2 Me m b e r 3 X Z Snow - Loading (kips) 20 20 Me m b e r 1 Member2 Me m b e r 3 X Z Load combination factors Load combination Se l f W e i g h t De a d Li v e Wi n d Sn o w 1.0D + 1.0L (Strength) 1.00 1.00 1.00 1.0D + 0.75L + 0.75S + 0.45W (Strength) 1.00 1.00 0.75 0.45 0.6D + 0.6W (Strength) 0.60 0.60 0.60 1.0D + 1.0S (Strength) 1.00 1.00 1.00 1.0D + 0.75L + 0.75S (Strength) 1.00 1.00 0.75 0.75 Node loads Project Job Ref. Section New Column/Beam Frame Sheet no./rev. 4 Calc. by N Date 3/27/2020 Chk'd by Date App'd by Date Node Load case Force Moment X Z (kips) (kips) (kip_ft) 2 Dead 0 7 0 3 Dead 0 7 0 2 Wind 14 0 0 2 Snow 0 20 0 3 Snow 0 20 0 Member Loads Member Load case Load Type Orientation Description Member2 Dead UDL GlobalZ 1.5 kips/ft Member2 Live UDL GlobalZ 2.1 kips/ft Results Forces Strength combinations - Moment envelope (kip_ft) 29.5 -60.7 250 -68.7 43.2 -68.7 Strength combinations - Shear envelope (kips) 2.1 -8.2 47.8 -47.8 10.2 3.5 Project Job Ref. Section New Column/Beam Frame Sheet no./rev. 5 Calc. by N Date 3/27/2020 Chk'd by Date App'd by Date ; Resistance factors Shear;v = 0.90 Flexure;b = 0.90 Tensile yielding;t,y = 0.90 Tensile rupture;t,r = 0.75 Compression;c = 0.90 Member1 design Section details Section type;HSS 8x8x3/8 (AISC 15th Edn (v15.0)) ASTM steel designation;A500 Gr.B Steel yield stress;Fy = 46 ksi Steel tensile stress;Fu = 58 ksi Modulus of elasticity;E = 29000 ksi 8" 0.35" 8" HSS 8x8x3/8 (AISC 15th Edn (v15.0)) Section depth, d, 8 in Section breadth, bf, 8 in Weight of section, Weight, 37.7 lbf/ft Section thickness, t, 0.349 in Area of section, A, 10.4 in2 Radius of gyration about x-axis, rx, 3.1 in Radius of gyration about y-axis, ry, 3.1 in Elastic section modulus about x-axis, Sx, 24.9 in3 Elastic section modulus about y-axis, Sy, 24.9 in3 Plastic section modulus about x-axis, Zx, 29.4 in3 Plastic section modulus about y-axis, Zy, 29.4 in3 Second moment of area about x-axis, Ix, 100 in4 Second moment of area about y-axis, Iy, 100 in4 Lateral restraint Both flanges have lateral restraint at supports only Consider Combination 1 - 1.0D + 1.0L (Strength) Classification of sections for local buckling - Section B4 Classification of flanges in flexure - Table B4.1b (case 17) Width to thickness ratio;(bf - 3  t) / t = 19.92 Limiting ratio for compact section;pff = 1.12  [E / Fy] = 28.12 Limiting ratio for non-compact section;rff = 1.40  [E / Fy] = 35.15;Compact Classification of web in flexure - Table B4.1b (case 19) Width to thickness ratio;(d - 3  t) / t = 19.92 Limiting ratio for compact section;pwf = 2.42  [E / Fy] = 60.76 Limiting ratio for non-compact section;rwf = 5.70  [E / Fy] = 143.12;Compact Section is compact in flexure Project Job Ref. Section New Column/Beam Frame Sheet no./rev. 6 Calc. by N Date 3/27/2020 Chk'd by Date App'd by Date Classification of flanges in uniform compression - Table B4.1a (case 6) Width to thickness ratio;(bf - 3  t) / t = 19.92 Limiting ratio for non-compact section;rfc = 1.40  [E / Fy] = 35.15;Nonslender Classification of web in uniform compression - Table B4.1a (case 6) Width to thickness ratio;(d - 3  t) / t = 19.92 Limiting ratio for non-compact section;rwc = 1.40  [E / Fy] = 35.15;Nonslender Section is nonslender in compression Check design at start of span Design of members for compression - Chapter E Required compressive strength;Pr = 55.2 kips Slenderness limitations and effective length - Section E2 Unbraced length;Lb,x = Lm1_s1 = 11 ft Effective length factor;Kx = 1.00 Column slenderness;x = Kx  Lb,x / rx = 42.581 Major axis column slenderness ratio does not exceed recommended limit of 200 Slenderness limitations and effective length - Section E2 Unbraced length;Lb,y = Lm1_s1_seg1 = 11 ft Effective length factor;Ky = 1.00 Column slenderness;y = Ky  Lb,y / ry = 42.581 Minor axis column slenderness ratio does not exceed recommended limit of 200 Flexural buckling of members without slender elements - Section E3 Elastic critical buckling stress - eq E3-4;Fe,x = 2  E / (Kx  Lb,x / rx)2 = 157.9 ksi Flexural buckling stress - eq E3-2;Fcr,x = [0.658Fy / Fe,x]  Fy = 40.7 ksi Nominal compressive strength for flexural buckling - eq E3-1; Pn,fb,x = Fcr,x  A = 423.5 kips Elastic critical buckling stress - eq E3-4;Fe,y = 2  E / (Ky  Lb,y / ry)2 = 157.9 ksi Flexural buckling stress - eq E3-2;Fcr,y = [0.658Fy / Fe,y]  Fy = 40.7 ksi Nominal compressive strength for flexural buckling - eq E3-1; Pn,fb,y = Fcr,y  A = 423.5 kips Design compressive strength - E1 Nominal compressive strength;Pn = min(Pn,fb,x, Pn,fb,y) = 423.5 kips Design compressive strength;Pc = c  Pn = 381.1 kips Pr / Pc = 0.145 PASS - Nominal compressive strength exceeds required compressive strength Design of members for shear - Chapter G Required shear strength;Vr,x = 8.2 kips Web area;Aw = 2  (d - 3  t)  t = 4.853 in2 Web plate buckling coefficient;kv = 5 (d - 3  t) / t <= 1.10  (kv  E / Fy) Web shear coefficient - eq G2-3;Cv = 1.000 Nominal shear strength - eq G2-1;Vn,x = 0.6  Fy  Aw  Cv = 133.9 kips Project Job Ref. Section New Column/Beam Frame Sheet no./rev. 7 Calc. by N Date 3/27/2020 Chk'd by Date App'd by Date Resistance factor;v = 0.90 Design shear strength;Vc,x = v  Vn,x = 120.6 kips Vr,x / Vc,x = 0.068 PASS - Design shear strength exceeds required shear strength Design of members for flexure - Chapter F Required flexural strength;Mr,x = 29.5 kips_ft Yielding - Section F7.1 Nominal flexural strength for yielding - eq F7-1;Mn,yld,x = Mp,x = Fy  Zx = 112.7 kips_ft Design flexural strength - F1 Nominal flexural strength;Mn,x = Mn,yld,x = 112.7 kips_ft Design flexural strength;Mc,x = b  Mn,x = 101.4 kips_ft Mr,x / Mc,x = 0.291 PASS - Design flexural strength exceeds required flexural strength Design of members for combined forces - Chapter H Combined flexure and axial force - eq H1-1b;Pr / (2  Pc) + Mr,x / Mc,x = 0.363 PASS - Combined flexure and axial force is within acceptable limits Check design at end of span Design of members for compression - Chapter E Required compressive strength;Pr = 54.8 kips Slenderness limitations and effective length - Section E2 Unbraced length;Lb,x = Lm1_s1 = 11 ft Effective length factor;Kx = 1.00 Column slenderness;x = Kx  Lb,x / rx = 42.581 Major axis column slenderness ratio does not exceed recommended limit of 200 Slenderness limitations and effective length - Section E2 Unbraced length;Lb,y = Lm1_s1_seg1 = 11 ft Effective length factor;Ky = 1.00 Column slenderness;y = Ky  Lb,y / ry = 42.581 Minor axis column slenderness ratio does not exceed recommended limit of 200 Flexural buckling of members without slender elements - Section E3 Elastic critical buckling stress - eq E3-4;Fe,x = 2  E / (Kx  Lb,x / rx)2 = 157.9 ksi Flexural buckling stress - eq E3-2;Fcr,x = [0.658Fy / Fe,x]  Fy = 40.7 ksi Nominal compressive strength for flexural buckling - eq E3-1; Pn,fb,x = Fcr,x  A = 423.5 kips Elastic critical buckling stress - eq E3-4;Fe,y = 2  E / (Ky  Lb,y / ry)2 = 157.9 ksi Flexural buckling stress - eq E3-2;Fcr,y = [0.658Fy / Fe,y]  Fy = 40.7 ksi Nominal compressive strength for flexural buckling - eq E3-1; Pn,fb,y = Fcr,y  A = 423.5 kips Design compressive strength - E1 Nominal compressive strength;Pn = min(Pn,fb,x, Pn,fb,y) = 423.5 kips Design compressive strength;Pc = c  Pn = 381.1 kips Project Job Ref. Section New Column/Beam Frame Sheet no./rev. 8 Calc. by N Date 3/27/2020 Chk'd by Date App'd by Date Pr / Pc = 0.144 PASS - Nominal compressive strength exceeds required compressive strength Design of members for shear - Chapter G Required shear strength;Vr,x = 8.2 kips Web area;Aw = 2  (d - 3  t)  t = 4.853 in2 Web plate buckling coefficient;kv = 5 (d - 3  t) / t <= 1.10  (kv  E / Fy) Web shear coefficient - eq G2-3;Cv = 1.000 Nominal shear strength - eq G2-1;Vn,x = 0.6  Fy  Aw  Cv = 133.9 kips Resistance factor;v = 0.90 Design shear strength;Vc,x = v  Vn,x = 120.6 kips Vr,x / Vc,x = 0.068 PASS - Design shear strength exceeds required shear strength Design of members for flexure - Chapter F Required flexural strength;Mr,x = 60.7 kips_ft Yielding - Section F7.1 Nominal flexural strength for yielding - eq F7-1;Mn,yld,x = Mp,x = Fy  Zx = 112.7 kips_ft Design flexural strength - F1 Nominal flexural strength;Mn,x = Mn,yld,x = 112.7 kips_ft Design flexural strength;Mc,x = b  Mn,x = 101.4 kips_ft Mr,x / Mc,x = 0.598 PASS - Design flexural strength exceeds required flexural strength Design of members for combined forces - Chapter H Combined flexure and axial force - eq H1-1b;Pr / (2  Pc) + Mr,x / Mc,x = 0.670 PASS - Combined flexure and axial force is within acceptable limits Consider Combination 3 - 0.6D + 0.6W (Strength) Check design at end of span Design of members for x-x axis deflection Maximum deflection;x = 0.319 in Allowable deflection;x,Allowable = Lm1_s1 / 240 = 0.55 in x / x,Allowable = 0.58 PASS - Allowable deflection exceeds design deflection Member2 design Section details Section type;W 16x77 (AISC 15th Edn (v15.0)) ASTM steel designation;A992 Steel yield stress;Fy = 50 ksi Steel tensile stress;Fu = 65 ksi Modulus of elasticity;E = 29000 ksi Project Job Ref. Section New Column/Beam Frame Sheet no./rev. 9 Calc. by N Date 3/27/2020 Chk'd by Date App'd by Date 10.3" 0.46" 16 . 5 " 0 . 7 6 " 0 . 7 6 " W 16x77 (AISC 15th Edn (v15.0)) Section depth, d, 16.5 in Section breadth, bf, 10.3 in Weight of section, Weight, 77 lbf/ft Flange thickness, tf, 0.76 in Web thickness, tw, 0.455 in Area of section, A, 22.6 in 2 Radius of gyration about x-axis, r x, 7 in Radius of gyration about y-axis, r y, 2.47 in Elastic section modulus about x-axis, Sx, 134 in3 Elastic section modulus about y-axis, Sy, 26.9 in3 Plastic section modulus about x-axis, Zx, 150 in3 Plastic section modulus about y-axis, Zy, 41.1 in3 Second moment of area about x-axis, I x, 1110 in 4 Second moment of area about y-axis, I y, 138 in4 Lateral restraint Both flanges have lateral restraint at supports only Consider Combination 1 - 1.0D + 1.0L (Strength) Classification of sections for local buckling - Section B4 Classification of flanges in flexure - Table B4.1b (case 10) Width to thickness ratio;bf / (2  tf) = 6.78 Limiting ratio for compact section;pff = 0.38  [E / Fy] = 9.15 Limiting ratio for non-compact section;rff = 1.0  [E / Fy] = 24.08;Compact Classification of web in flexure - Table B4.1b (case 15) Width to thickness ratio;(d - 2  k) / tw = 31.16 Limiting ratio for compact section;pwf = 3.76  [E / Fy] = 90.55 Limiting ratio for non-compact section;rwf = 5.70  [E / Fy] = 137.27;Compact Section is compact in flexure Classification of flanges in uniform compression - Table B4.1a (case 1) Width to thickness ratio;bf / (2  tf) = 6.78 Limiting ratio for non-compact section;rfc = 0.56  [E / Fy] = 13.49;Nonslender Classification of web in uniform compression - Table B4.1a (case 5) Width to thickness ratio;(d - 2  k) / tw = 31.16 Limiting ratio for non-compact section;rwc = 1.49  [E / Fy] = 35.88;Nonslender Section is nonslender in compression Check design at start of span Design of members for compression - Chapter E Required compressive strength;Pr = 8.2 kips Slenderness limitations and effective length - Section E2 Unbraced length;Lb,x = Lm2_s1 = 26 ft Effective length factor;Kx = 1.00 Project Job Ref. Section New Column/Beam Frame Sheet no./rev. 10 Calc. by N Date 3/27/2020 Chk'd by Date App'd by Date Column slenderness;x = Kx  Lb,x / rx = 44.571 Major axis column slenderness ratio does not exceed recommended limit of 200 Slenderness limitations and effective length - Section E2 Unbraced length;Lb,y = Lm2_s1_seg1 = 26 ft Effective length factor;Ky = 1.00 Column slenderness;y = Ky  Lb,y / ry = 126.316 Minor axis column slenderness ratio does not exceed recommended limit of 200 Flexural buckling of members without slender elements - Section E3 Elastic critical buckling stress - eq E3-4;Fe,x = 2  E / (Kx  Lb,x / rx)2 = 144.1 ksi Flexural buckling stress - eq E3-2;Fcr,x = [0.658Fy / Fe,x]  Fy = 43.2 ksi Nominal compressive strength for flexural buckling - eq E3-1; Pn,fb,x = Fcr,x  A = 977.2 kips Elastic critical buckling stress - eq E3-4;Fe,y = 2  E / (Ky  Lb,y / ry)2 = 17.9 ksi Flexural buckling stress - eq E3-3;Fcr,y = 0.877  Fe,y = 15.7 ksi Nominal compressive strength for flexural buckling - eq E3-1; Pn,fb,y = Fcr,y  A = 355.5 kips Torsional and torsional-flexural buckling of members without slender elements - Section E4 Unbraced length;Lb,z = Lm2_s1_seg1_R = 26 ft Effective length factor;Kz = 1.00 Flexural-torsional elastic buckling stress - eq E4-4;Fe = (2  E  Cw / (Kz  Lb,z)2 + G  J) / (Ix + Iy) = 52.1 ksi Flexural-torsional buckling stress - eq E3-2;Fcr = [0.658Fy / Fe]  Fy = 33.5 ksi Nominal compressive strength for torsional and flexural-torsional buckling - eq E4-1 ; Pn,ftb = Fcr  A = 756.4 kips Allowable compressive strength - E1 Nominal compressive strength;Pn = min(Pn,fb,x, Pn,fb,y, Pn,ftb) = 355.5 kips Allowable compressive strength;Pc = Pn / c = 212.9 kips Pr / Pc = 0.038 PASS - Nominal compressive strength exceeds required compressive strength Design of members for shear - Chapter G Required shear strength;Vr,x = 47.8 kips Web area;Aw = d  tw = 7.507 in2 Web plate buckling coefficient;kv = 5 (d - 2  k) / tw <= 2.24  (E / Fy) Web shear coefficient - eq G2-2;Cv = 1.000 Nominal shear strength - eq G2-1;Vn,x = 0.6  Fy  Aw  Cv = 225.2 kips Safety factor;v = 1.50 Allowable shear strength;Vc,x = Vn,x / v = 150.2 kips Vr,x / Vc,x = 0.318 PASS - Allowable shear strength exceeds required shear strength Design of members for flexure - Chapter F Required flexural strength;Mr,x = 60.7 kips_ft Project Job Ref. Section New Column/Beam Frame Sheet no./rev. 11 Calc. by N Date 3/27/2020 Chk'd by Date App'd by Date Yielding - Section F2.1 Nominal flexural strength for yielding - eq F2-1;Mn,yld,x = Mp,x = Fy  Zx = 625 kips_ft Lateral-torsional buckling - Section F2.2 Unbraced length;Lb = 26 ft Limiting unbraced length for yielding - eq F2-5;Lp = 1.76  ry  (E / Fy) = 8.725 ft Distance between flange centroids ;ho = 15.7 in c = 1 rts = 2.85 in Limiting unbraced length for inelastic LTB - eq F2-6;Lr = 1.95  rts  E / (0.7  Fy)  ((J  c / (Sx  ho)) + ((J  c / (Sx  ho))2 + 6.76  (0.7  Fy / E)2)) = 27.842 ft Moment at quarter point of segment;MA = 172.4 kips_ft Moment at center-line of segment;MB = 250 kips_ft Moment at three quarter point of segment ;MC = 172.4 kips_ft Maximum moment in segment;Mmax = 250 kips_ft LTB modification factor - eq F1-1;Cb = 12.5  Mmax / (2.5  Mmax + 3  MA + 4  MB + 3  MC) = 1.175 Nominal flexural strength for lateral-torsional buckling - eq F2-2 Mn,ltb,x = min(Cb  (Mp,x - (Mp,x - 0.7  Fy  Sx)  (Lb - Lp) / (Lr - Lp)), Mp,x) = 485.8 kips_ft Allowable flexural strength - F1 Nominal flexural strength;Mn,x = min(Mn,yld,x, Mn,ltb,x) = 485.8 kips_ft Allowable flexural strength;Mc,x = Mn,x / b = 290.9 kips_ft Mr,x / Mc,x = 0.209 PASS - Allowable flexural strength exceeds required flexural strength Design of members for combined forces - Chapter H Combined flexure and axial force - eq H1-1b;Pr / (2  Pc) + Mr,x / Mc,x = 0.228 PASS - Combined flexure and axial force is within acceptable limits Check design at start of span Design of members for compression - Chapter E Required compressive strength;Pr = 8.2 kips Slenderness limitations and effective length - Section E2 Unbraced length;Lb,x = Lm2_s1 = 26 ft Effective length factor;Kx = 1.00 Column slenderness;x = Kx  Lb,x / rx = 44.571 Major axis column slenderness ratio does not exceed recommended limit of 200 Slenderness limitations and effective length - Section E2 Unbraced length;Lb,y = Lm2_s1_seg1 = 26 ft Effective length factor;Ky = 1.00 Column slenderness;y = Ky  Lb,y / ry = 126.316 Minor axis column slenderness ratio does not exceed recommended limit of 200 Flexural buckling of members without slender elements - Section E3 Elastic critical buckling stress - eq E3-4;Fe,x = 2  E / (Kx  Lb,x / rx)2 = 144.1 ksi Flexural buckling stress - eq E3-2;Fcr,x = [0.658Fy / Fe,x]  Fy = 43.2 ksi Project Job Ref. Section New Column/Beam Frame Sheet no./rev. 12 Calc. by N Date 3/27/2020 Chk'd by Date App'd by Date Nominal compressive strength for flexural buckling - eq E3-1; Pn,fb,x = Fcr,x  A = 977.2 kips Elastic critical buckling stress - eq E3-4;Fe,y = 2  E / (Ky  Lb,y / ry)2 = 17.9 ksi Flexural buckling stress - eq E3-3;Fcr,y = 0.877  Fe,y = 15.7 ksi Nominal compressive strength for flexural buckling - eq E3-1; Pn,fb,y = Fcr,y  A = 355.5 kips Torsional and torsional-flexural buckling of members without slender elements - Section E4 Unbraced length;Lb,z = Lm2_s1_seg1_R = 26 ft Effective length factor;Kz = 1.00 Flexural-torsional elastic buckling stress - eq E4-4;Fe = (2  E  Cw / (Kz  Lb,z)2 + G  J) / (Ix + Iy) = 52.1 ksi Flexural-torsional buckling stress - eq E3-2;Fcr = [0.658Fy / Fe]  Fy = 33.5 ksi Nominal compressive strength for torsional and flexural-torsional buckling - eq E4-1 ; Pn,ftb = Fcr  A = 756.4 kips Allowable compressive strength - E1 Nominal compressive strength;Pn = min(Pn,fb,x, Pn,fb,y, Pn,ftb) = 355.5 kips Allowable compressive strength;Pc = Pn / c = 212.9 kips Pr / Pc = 0.038 PASS - Nominal compressive strength exceeds required compressive strength Design of members for flexure - Chapter F Required flexural strength;Mr,x = 250 kips_ft Yielding - Section F2.1 Nominal flexural strength for yielding - eq F2-1;Mn,yld,x = Mp,x = Fy  Zx = 625 kips_ft Lateral-torsional buckling - Section F2.2 Unbraced length;Lb = 26 ft Limiting unbraced length for yielding - eq F2-5;Lp = 1.76  ry  (E / Fy) = 8.725 ft Distance between flange centroids ;ho = 15.7 in c = 1 rts = 2.85 in Limiting unbraced length for inelastic LTB - eq F2-6;Lr = 1.95  rts  E / (0.7  Fy)  ((J  c / (Sx  ho)) + ((J  c / (Sx  ho))2 + 6.76  (0.7  Fy / E)2)) = 27.842 ft Moment at quarter point of segment;MA = 172.4 kips_ft Moment at center-line of segment;MB = 250 kips_ft Moment at three quarter point of segment ;MC = 172.4 kips_ft Maximum moment in segment;Mmax = 250 kips_ft LTB modification factor - eq F1-1;Cb = 12.5  Mmax / (2.5  Mmax + 3  MA + 4  MB + 3  MC) = 1.175 Nominal flexural strength for lateral-torsional buckling - eq F2-2 Mn,ltb,x = min(Cb  (Mp,x - (Mp,x - 0.7  Fy  Sx)  (Lb - Lp) / (Lr - Lp)), Mp,x) = 485.8 kips_ft Allowable flexural strength - F1 Nominal flexural strength;Mn,x = min(Mn,yld,x, Mn,ltb,x) = 485.8 kips_ft Allowable flexural strength;Mc,x = Mn,x / b = 290.9 kips_ft Mr,x / Mc,x = 0.859 PASS - Allowable flexural strength exceeds required flexural strength Project Job Ref. Section New Column/Beam Frame Sheet no./rev. 13 Calc. by N Date 3/27/2020 Chk'd by Date App'd by Date Design of members for combined forces - Chapter H Combined flexure and axial force - eq H1-1b;Pr / (2  Pc) + Mr,x / Mc,x = 0.879 PASS - Combined flexure and axial force is within acceptable limits Check design at start of span Design of members for compression - Chapter E Required compressive strength;Pr = 8.2 kips Slenderness limitations and effective length - Section E2 Unbraced length;Lb,x = Lm2_s1 = 26 ft Effective length factor;Kx = 1.00 Column slenderness;x = Kx  Lb,x / rx = 44.571 Major axis column slenderness ratio does not exceed recommended limit of 200 Slenderness limitations and effective length - Section E2 Unbraced length;Lb,y = Lm2_s1_seg1 = 26 ft Effective length factor;Ky = 1.00 Column slenderness;y = Ky  Lb,y / ry = 126.316 Minor axis column slenderness ratio does not exceed recommended limit of 200 Flexural buckling of members without slender elements - Section E3 Elastic critical buckling stress - eq E3-4;Fe,x = 2  E / (Kx  Lb,x / rx)2 = 144.1 ksi Flexural buckling stress - eq E3-2;Fcr,x = [0.658Fy / Fe,x]  Fy = 43.2 ksi Nominal compressive strength for flexural buckling - eq E3-1; Pn,fb,x = Fcr,x  A = 977.2 kips Elastic critical buckling stress - eq E3-4;Fe,y = 2  E / (Ky  Lb,y / ry)2 = 17.9 ksi Flexural buckling stress - eq E3-3;Fcr,y = 0.877  Fe,y = 15.7 ksi Nominal compressive strength for flexural buckling - eq E3-1; Pn,fb,y = Fcr,y  A = 355.5 kips Torsional and torsional-flexural buckling of members without slender elements - Section E4 Unbraced length;Lb,z = Lm2_s1_seg1_R = 26 ft Effective length factor;Kz = 1.00 Flexural-torsional elastic buckling stress - eq E4-4;Fe = (2  E  Cw / (Kz  Lb,z)2 + G  J) / (Ix + Iy) = 52.1 ksi Flexural-torsional buckling stress - eq E3-2;Fcr = [0.658Fy / Fe]  Fy = 33.5 ksi Nominal compressive strength for torsional and flexural-torsional buckling - eq E4-1 ; Pn,ftb = Fcr  A = 756.4 kips Allowable compressive strength - E1 Nominal compressive strength;Pn = min(Pn,fb,x, Pn,fb,y, Pn,ftb) = 355.5 kips Allowable compressive strength;Pc = Pn / c = 212.9 kips Pr / Pc = 0.038 PASS - Nominal compressive strength exceeds required compressive strength Design of members for shear - Chapter G Required shear strength;Vr,x = 47.8 kips Web area;Aw = d  tw = 7.507 in2 Web plate buckling coefficient;kv = 5 (d - 2  k) / tw <= 2.24  (E / Fy) Project Job Ref. Section New Column/Beam Frame Sheet no./rev. 14 Calc. by N Date 3/27/2020 Chk'd by Date App'd by Date Web shear coefficient - eq G2-2;Cv = 1.000 Nominal shear strength - eq G2-1;Vn,x = 0.6  Fy  Aw  Cv = 225.2 kips Safety factor;v = 1.50 Allowable shear strength;Vc,x = Vn,x / v = 150.2 kips Vr,x / Vc,x = 0.318 PASS - Allowable shear strength exceeds required shear strength Design of members for flexure - Chapter F Required flexural strength;Mr,x = 60.7 kips_ft Yielding - Section F2.1 Nominal flexural strength for yielding - eq F2-1;Mn,yld,x = Mp,x = Fy  Zx = 625 kips_ft Lateral-torsional buckling - Section F2.2 Unbraced length;Lb = 26 ft Limiting unbraced length for yielding - eq F2-5;Lp = 1.76  ry  (E / Fy) = 8.725 ft Distance between flange centroids ;ho = 15.7 in c = 1 rts = 2.85 in Limiting unbraced length for inelastic LTB - eq F2-6;Lr = 1.95  rts  E / (0.7  Fy)  ((J  c / (Sx  ho)) + ((J  c / (Sx  ho))2 + 6.76  (0.7  Fy / E)2)) = 27.842 ft Moment at quarter point of segment;MA = 172.4 kips_ft Moment at center-line of segment;MB = 250 kips_ft Moment at three quarter point of segment ;MC = 172.4 kips_ft Maximum moment in segment;Mmax = 250 kips_ft LTB modification factor - eq F1-1;Cb = 12.5  Mmax / (2.5  Mmax + 3  MA + 4  MB + 3  MC) = 1.175 Nominal flexural strength for lateral-torsional buckling - eq F2-2 Mn,ltb,x = min(Cb  (Mp,x - (Mp,x - 0.7  Fy  Sx)  (Lb - Lp) / (Lr - Lp)), Mp,x) = 485.8 kips_ft Allowable flexural strength - F1 Nominal flexural strength;Mn,x = min(Mn,yld,x, Mn,ltb,x) = 485.8 kips_ft Allowable flexural strength;Mc,x = Mn,x / b = 290.9 kips_ft Mr,x / Mc,x = 0.209 PASS - Allowable flexural strength exceeds required flexural strength Design of members for combined forces - Chapter H Combined flexure and axial force - eq H1-1b;Pr / (2  Pc) + Mr,x / Mc,x = 0.228 PASS - Combined flexure and axial force is within acceptable limits Check design 13ft along span Design of members for x-x axis deflection Maximum deflection;x = 0.968 in Allowable deflection;x,Allowable = Lm2_s1 / 240 = 1.3 in x / x,Allowable = 0.744 PASS - Allowable deflection exceeds design deflection Project Job Ref. Section New Column/Beam Frame Sheet no./rev. 15 Calc. by N Date 3/27/2020 Chk'd by Date App'd by Date Member3 design Section details Section type;HSS 8x8x3/8 (AISC 15th Edn (v15.0)) ASTM steel designation;A500 Gr.B Steel yield stress;Fy = 46 ksi Steel tensile stress;Fu = 58 ksi Modulus of elasticity;E = 29000 ksi 8" 0.35" 8" HSS 8x8x3/8 (AISC 15th Edn (v15.0)) Section depth, d, 8 in Section breadth, bf, 8 in Weight of section, Weight, 37.7 lbf/ft Section thickness, t, 0.349 in Area of section, A, 10.4 in2 Radius of gyration about x-axis, rx, 3.1 in Radius of gyration about y-axis, ry, 3.1 in Elastic section modulus about x-axis, Sx, 24.9 in3 Elastic section modulus about y-axis, Sy, 24.9 in3 Plastic section modulus about x-axis, Zx, 29.4 in3 Plastic section modulus about y-axis, Zy, 29.4 in3 Second moment of area about x-axis, Ix, 100 in4 Second moment of area about y-axis, Iy, 100 in4 Lateral restraint Both flanges have lateral restraint at supports only Consider Combination 2 - 1.0D + 0.75L + 0.75S + 0.45W (Strength) Classification of sections for local buckling - Section B4 Classification of flanges in flexure - Table B4.1b (case 17) Width to thickness ratio;(bf - 3  t) / t = 19.92 Limiting ratio for compact section;pff = 1.12  [E / Fy] = 28.12 Limiting ratio for non-compact section;rff = 1.40  [E / Fy] = 35.15;Compact Classification of web in flexure - Table B4.1b (case 19) Width to thickness ratio;(d - 3  t) / t = 19.92 Limiting ratio for compact section;pwf = 2.42  [E / Fy] = 60.76 Limiting ratio for non-compact section;rwf = 5.70  [E / Fy] = 143.12;Compact Section is compact in flexure Classification of flanges in uniform compression - Table B4.1a (case 6) Width to thickness ratio;(bf - 3  t) / t = 19.92 Limiting ratio for non-compact section;rfc = 1.40  [E / Fy] = 35.15;Nonslender Classification of web in uniform compression - Table B4.1a (case 6) Width to thickness ratio;(d - 3  t) / t = 19.92 Limiting ratio for non-compact section;rwc = 1.40  [E / Fy] = 35.15;Nonslender Section is nonslender in compression Project Job Ref. Section New Column/Beam Frame Sheet no./rev. 16 Calc. by N Date 3/27/2020 Chk'd by Date App'd by Date Check design at start of span Design of members for compression - Chapter E Required compressive strength;Pr = 49.3 kips Slenderness limitations and effective length - Section E2 Unbraced length;Lb,x = Lm3_s1 = 11 ft Effective length factor;Kx = 1.00 Column slenderness;x = Kx  Lb,x / rx = 42.581 Major axis column slenderness ratio does not exceed recommended limit of 200 Slenderness limitations and effective length - Section E2 Unbraced length;Lb,y = Lm3_s1_seg1 = 11 ft Effective length factor;Ky = 1.00 Column slenderness;y = Ky  Lb,y / ry = 42.581 Minor axis column slenderness ratio does not exceed recommended limit of 200 Flexural buckling of members without slender elements - Section E3 Elastic critical buckling stress - eq E3-4;Fe,x = 2  E / (Kx  Lb,x / rx)2 = 157.9 ksi Flexural buckling stress - eq E3-2;Fcr,x = [0.658Fy / Fe,x]  Fy = 40.7 ksi Nominal compressive strength for flexural buckling - eq E3-1; Pn,fb,x = Fcr,x  A = 423.5 kips Elastic critical buckling stress - eq E3-4;Fe,y = 2  E / (Ky  Lb,y / ry)2 = 157.9 ksi Flexural buckling stress - eq E3-2;Fcr,y = [0.658Fy / Fe,y]  Fy = 40.7 ksi Nominal compressive strength for flexural buckling - eq E3-1; Pn,fb,y = Fcr,y  A = 423.5 kips Design compressive strength - E1 Nominal compressive strength;Pn = min(Pn,fb,x, Pn,fb,y) = 423.5 kips Design compressive strength;Pc = c  Pn = 381.1 kips Pr / Pc = 0.129 PASS - Nominal compressive strength exceeds required compressive strength Design of members for shear - Chapter G Required shear strength;Vr,x = 10.2 kips Web area;Aw = 2  (d - 3  t)  t = 4.853 in2 Web plate buckling coefficient;kv = 5 (d - 3  t) / t <= 1.10  (kv  E / Fy) Web shear coefficient - eq G2-3;Cv = 1.000 Nominal shear strength - eq G2-1;Vn,x = 0.6  Fy  Aw  Cv = 133.9 kips Resistance factor;v = 0.90 Design shear strength;Vc,x = v  Vn,x = 120.6 kips Vr,x / Vc,x = 0.084 PASS - Design shear strength exceeds required shear strength Design of members for flexure - Chapter F Required flexural strength;Mr,x = 68.7 kips_ft Yielding - Section F7.1 Nominal flexural strength for yielding - eq F7-1;Mn,yld,x = Mp,x = Fy  Zx = 112.7 kips_ft Project Job Ref. Section New Column/Beam Frame Sheet no./rev. 17 Calc. by N Date 3/27/2020 Chk'd by Date App'd by Date Design flexural strength - F1 Nominal flexural strength;Mn,x = Mn,yld,x = 112.7 kips_ft Design flexural strength;Mc,x = b  Mn,x = 101.4 kips_ft Mr,x / Mc,x = 0.677 PASS - Design flexural strength exceeds required flexural strength Design of members for combined forces - Chapter H Combined flexure and axial force - eq H1-1b;Pr / (2  Pc) + Mr,x / Mc,x = 0.742 PASS - Combined flexure and axial force is within acceptable limits Check design at start of span Design of members for compression - Chapter E Required compressive strength;Pr = 49.6 kips Slenderness limitations and effective length - Section E2 Unbraced length;Lb,x = Lm3_s1 = 11 ft Effective length factor;Kx = 1.00 Column slenderness;x = Kx  Lb,x / rx = 42.581 Major axis column slenderness ratio does not exceed recommended limit of 200 Slenderness limitations and effective length - Section E2 Unbraced length;Lb,y = Lm3_s1_seg1 = 11 ft Effective length factor;Ky = 1.00 Column slenderness;y = Ky  Lb,y / ry = 42.581 Minor axis column slenderness ratio does not exceed recommended limit of 200 Flexural buckling of members without slender elements - Section E3 Elastic critical buckling stress - eq E3-4;Fe,x = 2  E / (Kx  Lb,x / rx)2 = 157.9 ksi Flexural buckling stress - eq E3-2;Fcr,x = [0.658Fy / Fe,x]  Fy = 40.7 ksi Nominal compressive strength for flexural buckling - eq E3-1; Pn,fb,x = Fcr,x  A = 423.5 kips Elastic critical buckling stress - eq E3-4;Fe,y = 2  E / (Ky  Lb,y / ry)2 = 157.9 ksi Flexural buckling stress - eq E3-2;Fcr,y = [0.658Fy / Fe,y]  Fy = 40.7 ksi Nominal compressive strength for flexural buckling - eq E3-1; Pn,fb,y = Fcr,y  A = 423.5 kips Design compressive strength - E1 Nominal compressive strength;Pn = min(Pn,fb,x, Pn,fb,y) = 423.5 kips Design compressive strength;Pc = c  Pn = 381.1 kips Pr / Pc = 0.130 PASS - Nominal compressive strength exceeds required compressive strength Design of members for shear - Chapter G Required shear strength;Vr,x = 10.2 kips Web area;Aw = 2  (d - 3  t)  t = 4.853 in2 Web plate buckling coefficient;kv = 5 (d - 3  t) / t <= 1.10  (kv  E / Fy) Web shear coefficient - eq G2-3;Cv = 1.000 Nominal shear strength - eq G2-1;Vn,x = 0.6  Fy  Aw  Cv = 133.9 kips Project Job Ref. Section New Column/Beam Frame Sheet no./rev. 18 Calc. by N Date 3/27/2020 Chk'd by Date App'd by Date Resistance factor;v = 0.90 Design shear strength;Vc,x = v  Vn,x = 120.6 kips Vr,x / Vc,x = 0.084 PASS - Design shear strength exceeds required shear strength Design of members for flexure - Chapter F Required flexural strength;Mr,x = 43.2 kips_ft Yielding - Section F7.1 Nominal flexural strength for yielding - eq F7-1;Mn,yld,x = Mp,x = Fy  Zx = 112.7 kips_ft Design flexural strength - F1 Nominal flexural strength;Mn,x = Mn,yld,x = 112.7 kips_ft Design flexural strength;Mc,x = b  Mn,x = 101.4 kips_ft Mr,x / Mc,x = 0.426 PASS - Design flexural strength exceeds required flexural strength Design of members for combined forces - Chapter H Combined flexure and axial force - eq H1-1b;Pr / (2  Pc) + Mr,x / Mc,x = 0.491 PASS - Combined flexure and axial force is within acceptable limits Check design 2ft 5.118in along span Design of members for x-x axis deflection Maximum deflection;x = 0.327 in Allowable deflection;x,Allowable = Lm3_s1 / 240 = 0.55 in x / x,Allowable = 0.595 PASS - Allowable deflection exceeds design deflection Project Job Ref. Section Column Base Plate Sheet no./rev. 1 Calc. by N Date 3/26/2020 Chk'd by Date App'd by Date COLUMN BASE PLATE DESIGN In accordance with AISC Steel Design Guide 1 and AISC 360-10 Tedds calculation version 2.1.02 1.5"11"1.5" 1. 5 " 11 . 0 " 1. 5 " Plan on baseplate Elevation on baseplate 0.45 ksi 83.0 kips 32.0 kip_in 8.0 kips HSS 8x8x3/8 Bolt diameter - 0.9" Bolt embedment - 15.0" Flange/base weld - 0.3" Web/base weld - 0.3" Design forces and moments Axial force;Pu = 83.0 kips; (Compression) Bending moment;Mu = 32.0 kip_in Shear force;Fv = 8.0 kips Eccentricity;e = ABS(Mu / Pu) = 0.386 in Anchor bolt to center of plate;f = N/2 - e1 = 5.500 in Column details Column section;HSS 8x8x3/8 Depth;d = 8.000 in Breadth;bf = 8.000 in Thickness;t = 0.349 in Baseplate details Depth;N = 14.000 in Breadth;B = 14.000 in Thickness;tp = 0.750 in Design strength;Fy = 36.0 ksi Foundation geometry Member thickness;ha = 36.000 in Dist center of baseplate to left edge foundation;xce1 = 12.000 in Dist center of baseplate to right edge foundation;xce2 = 12.000 in Dist center of baseplate to bot edge foundation;yce1 = 12.000 in Dist center of baseplate to top edge foundation;yce2 = 12.000 in Holding down bolt and anchor plate details Total number of bolts;Nbolt = 4 Bolt diameter;do = 0.875 in Bolt spacing;sbolt = 11.000 in Edge distance;e1 = 1.500 in Project Job Ref. Section Column Base Plate Sheet no./rev. 2 Calc. by N Date 3/26/2020 Chk'd by Date App'd by Date Minimum tensile strength, base plate ;Fy = 36 ksi Minimum tensile strength, column;FyCol = 50 ksi Compressive strength of concrete;f’c = 3 ksi Strength reduction factors Compression;c = 0.65 Flexure;b = 0.90 Weld shear;v = 0.75 Plate cantilever dimensions Area of base plate;A1 = B  N = 196.000 in2 Maximum area of supporting surface;A2 = (N + 2  lmin)  (B + 2  lmin) = 576.000 in2 Nominal strength of concrete under base plate;Pp = 0.85  f'c  A1  min((A2 / A1), 2) = 856.8 kips Bending line cantilever distance m;m = (N - 0.95  d) / 2 = 3.200 in Bending line cantilever distance n;n = (B - 0.95  bf) / 2 = 3.200 in Maximum bending line cantilever;l = max(m, n) = 3.200 in Check eccentricity Maximum bearing stress;fp,max = 0.85  f'c  c  min((A2 / A1), 2) = 2.84 ksi Maximum bearing pressure;qmax = fp,max  B = 39.78 kips/in Critical eccentricity;ecrit = N / 2 - Pu / (2  qmax) = 5.957 in e <= ecrit so loads can be resisted by bearing alone. Therefore consider as a small moment Bearing length;Y = N - 2e = 13.229 in Bearing pressure;q = Pu / Y = 6.3 kips/in PASS - Maximum allowable bearing stress exceeds actual Base plate yielding limit at bearing interface Required plate thickness;tp,req = (4  (fp  (l2 / 2)) / (b  Fy)) = 0.532 in PASS - Thickness of plate exceeds required thickness Flange weld Flange weld leg length;twf = 0.3125 in Tension capacity of flange;Ptf = bf  t  FyCol = 139.6 kips Force in tension flange;Ftf = Mu / (d - t) - Pu  (bf  t) / Acol = -18.1 kips Critical force in flange;Ff = min(Ptf, max(Ftf, 0kips)) = 0.0 kips Flange weld force per in;Rwf = Ff / bf = 0.0 kips/in Electrode classification number;FEXX = 70.0 ksi Design weld stress;Fnw = v  0.60  FEXX  (1.0 + 0.5  (sin(90deg))1.5) = 47.250ksi Design strength of weld per in;Rnf = Fnw  twf / (2) = 10.4 kips/in PASS - Available strength of flange weld exceeds force in flange weld Shear weld Shear web weld leg length;tww = 0.3125 in Shear web weld force per in;Rwl = Fv / (2  (d - 2  t)) = 0.548 kips/in Electrode classification number;FEXX = 70.0 ksi Design weld stress;Fnw = v  0.60  FEXX  (1.0 + 0.5  (sin(0deg))1.5) = 31.500ksi Design strength of weld per in;Rnl = Fnw  tww / (2) = 7.0 kips/in PASS - Available strength of shear weld exceeds force in shear weld Project Job Ref. Section Column Base Plate Sheet no./rev. 3 Calc. by N Date 3/26/2020 Chk'd by Date App'd by Date ANCHOR BOLT DESIGN In accordance with ACI318-11 Tedds calculation version 2.1.02 Anchor bolt geometry Type of anchor bolt;Cast-in headed end bolt anchor Diameter of anchor bolt;da = 0.875 in Number of bolts in x direction;Nboltx = 2 Number of bolts in y direction;Nbolty = 2 Total number of bolts;ntotal = (Nboltx  2) + (Nbolty - 2)  2 = 4 Total number of bolts in tension ;ntens = NboltN = 0 Spacing of bolts in x direction;sboltx = 11 in Spacing of bolts in y direction;sbolty = 11 in Number of threads per inch;nt = 9 Effective cross-sectional area of anchor;Ase =  / 4  (da - 0.9743 in / nt)2 = 0.462 in2 Embedded depth of each anchor bolt;hef = 15 in Material details Minimum yield strength of steel;fya = 36 ksi Nominal tensile strength of steel;futa = 58 ksi Compressive strength of concrete;f’c = 3 ksi Concrete modification factor; = 1.00 Modification factor for cast-in anchor concrete failure a = 1.0  = 1.00 Strength reduction factors Tension of steel element;t,s = 0.75 Shear of steel element;v,s = 0.65 Concrete tension;t,c = 0.65 Concrete shear;v,c = 0.70 Concrete tension for pullout;t,cB = 0.70 Concrete shear for pryout;v,cB = 0.70 Shear force applied to bolt group ;V = 8.00 kips Steel strength of anchor in shear (D.6.1) Built-up grout pads are used so nominal strength will be multiplied by 0.8 (D.6.1.3) Effective number of anchors in shear;NboltV = 4 Nom strength of anchor in shear;Vsa = 0.8  NboltV  0.6  Ase  futa = 51.42 kips Steel strength of anchor in shear;Vsa = v,s  Vsa = 33.42 kips PASS - Steel strength of anchor exceeds shear in bolts Project Job Ref. Section Column Base Plate Sheet no./rev. 4 Calc. by N Date 3/26/2020 Chk'd by Date App'd by Date Concrete breakout strength in shear perpendicular to edge - Case 2. All shear resisted by rear bolts (D.6.3) 6. 5 " 11 . 0 " 6. 5 " 6.5"11.0"6.5" Plan on foundation Concrete breakout - shear 8 kips 2' 3' 1 ' 3 " Section A-A Applied shear;Vapp = V = 8.00 kips Edge distance x for shear near corner;ca1 = 17.5 in Edge distance y for shear near corner;ca2 = min(yce1, yce2) - (((Nbolty - 1)/2)  sbolty) = 6.5 in Load bearing length of anchor;le = min(hef, 8  da) = 7 in Basic concrete breakout strength;Vb1 = 7  (le / da)0.2  (da)  a  (f'c  1psi)  (ca1)1.5 = 39.80 kips Vb2 = 9  a  (f'c  1psi  1 in)  (ca1)1.5 = 36.09 kips Basic concrete breakout strength;Vb = Min(Vb1, Vb2) = 36.09 kips Projected area of a single anchor;AVco = 4.5  ca12 = 1378.1 in2 Projected area of a group of anchors;AVc = 630 in2 Mod factor for edge effect;ed.V = 0.7 + 0.3  ca2 / (1.5  ca1) = 0.774 Eccentricity of loading;e’V = 0 in Modification factor of eccentric loading;ec,V = min(1, 1 / (1 + ((2  e'V) / (3  ca1)))) = 1.000 Modification factor for cracking;c,V = 1.000 Modification factor for edge distance;h,V = 1.0 = 1.000 Nominal concrete break out strength in shear;Vcbg = AVc / AVco  ec,V  ed,V  c,V  h,V  Vb = 12.77 kips Concrete break out strength in shear;Vcbg = v,c Vcbg = 8.94 kips PASS - Shear breakout perpendicular to edge strength exceeds shear in bolts Project Job Ref. Section Column Base Plate Sheet no./rev. 5 Calc. by N Date 3/26/2020 Chk'd by Date App'd by Date Concrete breakout strength in shear - Case 1. Half of shear resisted by front bolts (D.6.3) 6. 5 " 11 . 0 " 6. 5 " 6.5"11.0"6.5" Plan on foundation Concrete breakout - shear 8 kips 2' 3' 1 ' 3 " Section A-A Applied shear;Vapp = V / 2 = 4.00 kips Edge distance x for shear near corner;ca1 = 6.5 in Edge distance y for shear near corner;ca2 = min(yce1, yce2) - (((Nbolty - 1)/2)  sbolty) = 6.5 in Load bearing length of anchor;le = min(hef, 8  da) = 7 in Basic concrete breakout strength;Vb1 = 7  (le / da)0.2  (da)  a  (f'c  1psi)  (ca1)1.5 = 9.01 kips Vb2 = 9  a  (f'c  1psi  1 in)  (ca1)1.5 = 8.17 kips Basic concrete breakout strength;Vb = Min(Vb1, Vb2) = 8.17 kips Projected area of a single anchor;AVco = 4.5  ca12 = 190.1 in2 Projected area of a group of anchors;AVc = 234 in2 Mod factor for edge effect;ed.V = 0.7 + 0.3  ca2 / (1.5  ca1) = 0.900 Eccentricity of loading;e’V = 0 in Modification factor of eccentric loading;ec,V = min(1, 1 / (1 + ((2  e'V) / (3  ca1)))) = 1.000 Modification factor for cracking;c,V = 1.000 Modification factor for edge distance;h,V = 1.0 = 1.000 Nominal concrete break out strength in shear;Vcbg = AVc / AVco  ec,V  ed,V  c,V  h,V  Vb = 9.05 kips Concrete break out strength in shear;Vcbg = v,c Vcbg = 6.33 kips PASS - Shear breakout perpendicular to edge strength exceeds shear in bolts Project Job Ref. Section Column Base Plate Sheet no./rev. 6 Calc. by N Date 3/26/2020 Chk'd by Date App'd by Date Concrete breakout strength in shear parallel to edge - Case 2. All shear resisted by rear bolts ( parallel to edge - Case 2. All shear resisted by rear bolts) 6. 5 " 11 . 0 " 6 . 5 " 6.5"11.0"6.5" Plan on foundation Concrete breakout - side shear 8 kips 2' 3' 1 ' 3 " Section A-A Applied shear;Vapp = V = 8.00 kips Edge distance x for shear near corner;ca1,p = 17.5 in Edge distance y for shear near corner;ca2,p = min(xce1, xce2) - (((Nboltx - 1)/2)  sboltx) = 6.5 in Load bearing length of anchor;le = min(hef, 8  da) = 7 in Basic concrete breakout strength;Vb,p1 = 7  (le / da)0.2  (da)  a  (f'c  1psi)  (ca1,p)1.5 = 39.80 kips Vb,p2 = 9  a  (f'c  1psi 1in)  (ca1,p)1.5 = 36.09 kips Basic concrete breakout strength;Vb,p = Min(Vb,p1, Vb,p2) = 36.09 kips Projected area of a single anchor;AVco,p = 4.5  ca1,p2 = 1378.1 in2 Projected area of a group of anchors;AVc,p = 630 in2 Mod factor for edge effect;ed,V,p = 1.000 Eccentricity of loading;e’V,p = 0 in Modification factor of eccentric loading;ec,V,p = min(1, 1 / (1 + ((2  e'V,p) / (3  ca1,p)))) = 1.000 Modification factor for cracking;c,V = 1.000 Modification factor for edge distance;h,V,p = 1.0 = 1.000 Nominal concrete break out strength in shear;Vcbg,p = 2  AVc,p / AVco,p  ec,V,p  ed,V,p  c,V  h,V,p  Vb,p = 32.99 kips Concrete break out strength in shear;Vcbg,p = v,c Vcbg,p = 23.10 kips PASS - Shear breakout strength parallel to edge exceeds shear in bolts Project Job Ref. Section Column Base Plate Sheet no./rev. 7 Calc. by N Date 3/26/2020 Chk'd by Date App'd by Date Concrete breakout strength in shear parallel to edge - Case 1. Half of shear resisted by front bolts ( parallel to edge - Case 1. Half of shear resisted by front bolts) 6. 5 " 11 . 0 " 6 . 5 " 6.5"11.0"6.5" Plan on foundation Concrete breakout - side shear 8 kips 2' 3' 1 ' 3 " Section A-A Applied shear;Vapp = V / 2 = 4.00 kips Edge distance x for shear near corner;ca1,p = 6.5 in Edge distance y for shear near corner;ca2,p = min(xce1, xce2) - (((Nboltx - 1)/2)  sboltx) = 6.5 in Load bearing length of anchor;le = min(hef, 8  da) = 7 in Basic concrete breakout strength;Vb,p1 = 7  (le / da)0.2  (da)  a  (f'c  1psi)  (ca1,p)1.5 = 9.01 kips Vb,p2 = 9  a  (f'c  1psi 1in)  (ca1,p)1.5 = 8.17 kips Basic concrete breakout strength;Vb,p = Min(Vb,p1, Vb,p2) = 8.17 kips Projected area of a single anchor;AVco,p = 4.5  ca1,p2 = 190.1 in2 Projected area of a group of anchors;AVc,p = 234 in2 Mod factor for edge effect;ed,V,p = 1.000 Eccentricity of loading;e’V,p = 0 in Modification factor of eccentric loading;ec,V,p = min(1, 1 / (1 + ((2  e'V,p) / (3  ca1,p)))) = 1.000 Modification factor for cracking;c,V = 1.000 Modification factor for edge distance;h,V,p = 1.0 = 1.000 Nominal concrete break out strength in shear;Vcbg,p = 2  AVc,p / AVco,p  ec,V,p  ed,V,p  c,V  h,V,p  Vb,p = 20.11 kips Concrete break out strength in shear;Vcbg,p = v,c Vcbg,p = 14.08 kips PASS - Shear breakout strength parallel to edge exceeds shear in bolts Project Job Ref. Section Column Base Plate Sheet no./rev. 8 Calc. by N Date 3/26/2020 Chk'd by Date App'd by Date Pryout strength of anchor in shear (D.6.3) 6. 5 " 11 . 0 " 6. 5 " 6.5"11.0"6.5" Plan on foundation Concrete breakout - tension 8 kips 2' 3' 1 ' 3 " Section A-A The anchors are located at less than 1.5hef from 4 edges. Therefore the effective embedded depth has to be limited to 4.33" in accordance with D.5.2.3 Limiting embedded depth;hef,lim = 4.33 in Coeff for basic breakout strength in tension;kc = 24 Breakout strength for single anchor in tension;Nb = kc  a  (f'c  1 psi)  hef,lim1.5  1 in0.5 = 11.86 kips Projected area for groups of anchors;ANc = 312 in2 Projected area of a single anchor;ANco = 9  hef,lim2 = 169 in2 Min dist center of anchor to edge of concrete;ca,min = 6.5 in Mod factor for groups loaded eccentrically;ec,N = min(1 / (1 + ((2  e'N) / (3  hef,lim))), 1) = 1.000 Modification factor for edge effects ;ed,N = 1.0 = 1.000 Modification factor for no cracking at service loads;c,N = 1.000 Modification factor for cracked concrete;cp,N = 1.000 Nominal concrete breakout strength;Ncbg = ANc / ANco  ed,N  c,N  cp,N  Nb = 21.89 kips Concrete breakout strength;Ncbg = t,c  Ncbg = 14.23 kips Coefficient of pryout strength;kcp = 2.0 Nominal pryout strength of anchor in shear;Vcpg = kcp  Ncbg = 43.78 kips Pryout strength of anchor in shear;Vcpg = v,cB  Vcpg = 30.65 kips PASS - Pryout strength of anchor exceeds shear in bolts ; Project Job Ref. Section Footing at Column near Entry (revised) Sheet no./rev. 1 Calc. by N Date 3/26/2020 Chk'd by Date App'd by Date FOUNDATION ANALYSIS & DESIGN (ACI318) In accordance with ACI318-11 incorporating Errata as of August 8, 2014 Tedds calculation version 3.2.09 FOOTING ANALYSIS Length of foundation;Lx = 7 ft Width of foundation;Ly = 4.5 ft Foundation area;A = Lx  Ly = 31.5 ft2 Depth of foundation;h = 12 in Depth of soil over foundation;hsoil = 12 in Density of concrete;conc = 150.0 lb/ft3 1 2.457 ksf 2.457 ksf 2.457 ksf 2.457 ksf x y Column no.1 details Length of column;lx1 = 12.00 in Width of column;ly1 = 12.00 in position in x-axis;x1 = 42.00 in position in y-axis;y1 = 27.00 in Soil properties Gross allowable bearing pressure;qallow_Gross = 3 ksf Density of soil;soil = 120.0 lb/ft3 Angle of internal friction;b = 30.0 deg Design base friction angle;bb = 30.0 deg Coefficient of base friction;tan(bb) = 0.577 Self weight;Fswt = h  conc = 150 psf Project Job Ref. Section Footing at Column near Entry (revised) Sheet no./rev. 2 Calc. by N Date 3/26/2020 Chk'd by Date App'd by Date Soil weight;Fsoil = hsoil  soil = 120 psf Column no.1 loads Dead load in z;FDz1 = 28.0 kips Live load in z;FLz1 = 32.0 kips Snow load in z;FSz1 = 22.5 kips Wind load moment in x;MWx1 = 20.0 kip_ft Wind load moment in y;MWy1 = 32.0 kip_ft Footing analysis for soil and stability Load combinations per ASCE 7-10 1.0D (0.386) 1.0D + 1.0L (0.725) 1.0D + 1.0S (0.624) 1.0D + 0.75L + 0.75S (0.819) 1.0D + 0.6W (0.766) Combination 7 results: 1.0D + 0.75L + 0.75S Forces on foundation Force in z-axis;Fdz = D  A  (Fswt + Fsoil) + D  FDz1 + L  FLz1 + S  FSz1 = 77.4 kips Moments on foundation Moment in x-axis, about x is 0;Mdx = D  (A  (Fswt + Fsoil)  Lx / 2) + D  (FDz1  x1) + L  (FLz1  x1) + S  (FSz1  x1) = 270.8 kip_ft Moment in y-axis, about y is 0;Mdy = D  (A  (Fswt + Fsoil)  Ly / 2) + D  (FDz1  y1) + L  (FLz1  y1) + S  (FSz1  y1) = 174.1 kip_ft Uplift verification Vertical force;Fdz = 77.38 kips PASS - Foundation is not subject to uplift Bearing resistance Eccentricity of base reaction Eccentricity of base reaction in x-axis;edx = Mdx / Fdz - Lx / 2 = 0 in Eccentricity of base reaction in y-axis;edy = Mdy / Fdz - Ly / 2 = 0 in Pad base pressures q1 = Fdz  (1 - 6  edx / Lx - 6  edy / Ly) / (Lx  Ly) = 2.457 ksf q2 = Fdz  (1 - 6  edx / Lx + 6  edy / Ly) / (Lx  Ly) = 2.457 ksf q3 = Fdz  (1 + 6  edx / Lx - 6  edy / Ly) / (Lx  Ly) = 2.457 ksf q4 = Fdz  (1 + 6  edx / Lx + 6  edy / Ly) / (Lx  Ly) = 2.457 ksf Minimum base pressure;qmin = min(q1,q2,q3,q4) = 2.457 ksf Maximum base pressure;qmax = max(q1,q2,q3,q4) = 2.457 ksf Allowable bearing capacity Allowable bearing capacity;qallow = qallow_Gross = 3 ksf qmax / qallow = 0.819 PASS - Allowable bearing capacity exceeds design base pressure Project Job Ref. Section Footing at Column near Entry (revised) Sheet no./rev. 3 Calc. by N Date 3/26/2020 Chk'd by Date App'd by Date FOOTING DESIGN (ACI318) In accordance with ACI318-11 incorporating Errata as of August 8, 2014 Material details Compressive strength of concrete;f’c = 4000 psi Yield strength of reinforcement;fy = 60000 psi Cover to reinforcement;cnom = 3 in Concrete type;Normal weight Concrete modification factor; = 1.00 Column type;Concrete Analysis and design of concrete footing Load combinations per ASCE 7-10 1.4D (0.274) 1.2D + 1.6L + 0.5Lr (0.594) 1.2D + 1.6L + 0.5S (0.673) 1.2D + 1.0L + 1.6S (0.712) 1.2D + 1.6Lr + 0.5W (0.223) Combination 6 results: 1.2D + 1.0L + 1.6S Forces on foundation Ultimate force in z-axis;Fuz = D  A  (Fswt + Fsoil) + D  FDz1 + L  FLz1 + S  FSz1 = 111.8 kips Moments on foundation Ultimate moment in x-axis, about x is 0;Mux = D  (A  (Fswt + Fsoil)  Lx / 2) + D  (FDz1  x1) + L  (FLz1  x1) + S  (FSz1  x1) = 391.3 kip_ft Ultimate moment in y-axis, about y is 0;Muy = D  (A  (Fswt + Fsoil)  Ly / 2) + D  (FDz1  y1) + L  (FLz1  y1) + S  (FSz1  y1) = 251.6 kip_ft Eccentricity of base reaction Eccentricity of base reaction in x-axis;eux = Mux / Fuz - Lx / 2 = 0 in Eccentricity of base reaction in y-axis;euy = Muy / Fuz - Ly / 2 = 0 in Pad base pressures qu1 = Fuz  (1 - 6  eux / Lx - 6  euy / Ly) / (Lx  Ly) = 3.549 ksf qu2 = Fuz  (1 - 6  eux / Lx + 6  euy / Ly) / (Lx  Ly) = 3.549 ksf qu3 = Fuz  (1 + 6  eux / Lx - 6  euy / Ly) / (Lx  Ly) = 3.549 ksf qu4 = Fuz  (1 + 6  eux / Lx + 6  euy / Ly) / (Lx  Ly) = 3.549 ksf Minimum ultimate base pressure;qumin = min(qu1,qu2,qu3,qu4) = 3.549 ksf Maximum ultimate base pressure;qumax = max(qu1,qu2,qu3,qu4) = 3.549 ksf Shear diagram, x axis (kips) 50.8 0 0 -50.8 33.8 -33.8 Project Job Ref. Section Footing at Column near Entry (revised) Sheet no./rev. 4 Calc. by N Date 3/26/2020 Chk'd by Date App'd by Date Moment diagram, x axis (kip_ft) 88.9 0 88.9 0 65.3 Moment design, x direction, positive moment Ultimate bending moment;Mu.x.max = 65.314 kip_ft Tension reinforcement provided;6 No.5 bottom bars (9.4 in c/c) Area of tension reinforcement provided;Asx.bot.prov = 1.86 in2 Minimum area of reinforcement (10.5.4);As.min = 0.0018  Ly  h = 1.166 in2 PASS - Area of reinforcement provided exceeds minimum Maximum spacing of reinforcement (10.5.4);smax = min(3  h, 18 in) = 18 in PASS - Maximum permissible reinforcement spacing exceeds actual spacing Depth to tension reinforcement;d = h - cnom - x.bot / 2 = 8.688 in Depth of compression block;a = Asx.bot.prov  fy / (0.85  f'c  Ly) = 0.608 in Neutral axis factor;1 = 0.85 Depth to neutral axis;c = a / 1 = 0.715 in Strain in tensile reinforcement (10.3.5);t = 0.003  d / c - 0.003 = 0.03345 PASS - Tensile strain exceeds minimum required, 0.004 Nominal moment capacity;Mn = Asx.bot.prov  fy  (d - a / 2) = 77.967 kip_ft Flexural strength reduction factor;f = min(max(0.65 + (t - 0.002)  (250 / 3), 0.65), 0.9) = 0.900 Design moment capacity;Mn = f  Mn = 70.171 kip_ft Mu.x.max / Mn = 0.931 PASS - Design moment capacity exceeds ultimate moment load One-way shear design, x direction Ultimate shear force;Vu.x = 33.791 kips Depth to reinforcement;dv = h - cnom - x.bot / 2 = 8.688 in Shear strength reduction factor;v = 0.75 Nominal shear capacity (Eq. 11-3);Vn = 2    (f'c  1 psi)  Ly  dv = 59.34 kips Design shear capacity;Vn = v  Vn = 44.505 kips Vu.x / Vn = 0.759 PASS - Design shear capacity exceeds ultimate shear load Shear diagram, y axis (kips) 50.8 0 0 -50.8 24.3 -24.3 Project Job Ref. Section Footing at Column near Entry (revised) Sheet no./rev. 5 Calc. by N Date 3/26/2020 Chk'd by Date App'd by Date Moment diagram, y axis (kip_ft) 57.1 0 57.1 0 34.6 Moment design, y direction, positive moment Ultimate bending moment;Mu.y.max = 34.572 kip_ft Tension reinforcement provided;9 No.5 bottom bars (9.6 in c/c) Area of tension reinforcement provided;Asy.bot.prov = 2.79 in2 Minimum area of reinforcement (10.5.4);As.min = 0.0018  Lx  h = 1.814 in2 PASS - Area of reinforcement provided exceeds minimum Maximum spacing of reinforcement (10.5.4);smax = min(3  h, 18 in) = 18 in PASS - Maximum permissible reinforcement spacing exceeds actual spacing Depth to tension reinforcement;d = h - cnom - x.bot - y.bot / 2 = 8.062 in Depth of compression block;a = Asy.bot.prov  fy / (0.85  f'c  Lx) = 0.586 in Neutral axis factor;1 = 0.85 Depth to neutral axis;c = a / 1 = 0.690 in Strain in tensile reinforcement (10.3.5);t = 0.003  d / c - 0.003 = 0.03208 PASS - Tensile strain exceeds minimum required, 0.004 Nominal moment capacity;Mn = Asy.bot.prov  fy  (d - a / 2) = 108.384 kip_ft Flexural strength reduction factor;f = min(max(0.65 + (t - 0.002)  (250 / 3), 0.65), 0.9) = 0.900 Design moment capacity;Mn = f  Mn = 97.545 kip_ft Mu.y.max / Mn = 0.354 PASS - Design moment capacity exceeds ultimate moment load Footing geometry factor (15.4.4.2);f = Lx / Ly = 1.556 Area of reinf. req. for uniform distribution (CRSI);Asreq = (Mu.y.max / (f  fy  (d - a / 2)))  2  f / (f + 1) = 1.204 in2 PASS - Reinforcement can be distributed uniformly One-way shear design, y direction Ultimate shear force;Vu.y = 24.342 kips Depth to reinforcement;dv = h - cnom - x.bot - y.bot / 2 = 8.062 in Shear strength reduction factor;v = 0.75 Nominal shear capacity (Eq. 11-3);Vn = 2    (f'c  1 psi)  Lx  dv = 85.666 kips Design shear capacity;Vn = v  Vn = 64.25 kips Vu.y / Vn = 0.379 PASS - Design shear capacity exceeds ultimate shear load Two-way shear design at column 1 Depth to reinforcement;dv2 = 8.375 in Shear perimeter length (11.11.1.2);lxp = 20.375 in Shear perimeter width (11.11.1.2);lyp = 20.375 in Project Job Ref. Section Footing at Column near Entry (revised) Sheet no./rev. 6 Calc. by N Date 3/26/2020 Chk'd by Date App'd by Date Shear perimeter (11.11.1.2);bo = 2  (lx1 + dv2) + 2  (ly1 + dv2) = 81.500 in Shear area;Ap = lx,perim  ly,perim = 415.141 in2 Surcharge loaded area;Asur = Ap - lx1  ly1 = 271.141 in2 Ultimate bearing pressure at center of shear area;qup.avg = 3.549 ksf Ultimate shear load;Fup = D  FDz1 + L  FLz1 + S  FSz1 + D  Ap  Fswt + D  Asur  Fsoil - qup.avg  Ap = 92.157 kips Ultimate shear stress from vertical load;vug = max(Fup / (bo  dv2),0 psi) = 135.017 psi Column geometry factor (11.11.2.1); = ly1 / lx1 = 1.00 Column location factor (11.11.2.1);s =40 Concrete shear strength (11.11.2.1);vcpa = (2 + 4 / )    (f'c  1 psi) = 379.473 psi vcpb = (s  dv2 / bo + 2)    (f'c  1 psi) = 386.457 psi vcpc = 4    (f'c  1 psi) = 252.982 psi vcp = min(vcpa,vcpb,vcpc) = 252.982 psi Shear strength reduction factor;v = 0.75 Nominal shear stress capacity (Eq. 11-2);vn = vcp = 252.982 psi Design shear stress capacity (Eq. 11-1);vn = v  vn = 189.737 psi vug / vn = 0.712 PASS - Design shear stress capacity exceeds ultimate shear stress load 1 6 No.5 bottom bars (9.4 in c/c) 9 No.5 bottom bars (9.6 in c/c) Project Job Ref. Section Footing at Existing Column Sheet no./rev. 1 Calc. by N Date 3/27/2020 Chk'd by Date App'd by Date FOUNDATION ANALYSIS & DESIGN (ACI318) In accordance with ACI318-11 incorporating Errata as of August 8, 2014 Tedds calculation version 3.2.09 FOOTING ANALYSIS Length of foundation;Lx = 8 ft Width of foundation;Ly = 4 ft Foundation area;A = Lx  Ly = 32 ft2 Depth of foundation;h = 12 in Depth of soil over foundation;hsoil = 12 in Density of concrete;conc = 150.0 lb/ft3 1 2.992 ksf 2.992 ksf 2.992 ksf 2.992 ksf x y Column no.1 details Length of column;lx1 = 12.00 in Width of column;ly1 = 12.00 in position in x-axis;x1 = 48.00 in position in y-axis;y1 = 24.00 in Soil properties Gross allowable bearing pressure;qallow_Gross = 3 ksf Density of soil;soil = 120.0 lb/ft3 Angle of internal friction;b = 30.0 deg Design base friction angle;bb = 30.0 deg Coefficient of base friction;tan(bb) = 0.577 Self weight;Fswt = h  conc = 150 psf Project Job Ref. Section Footing at Existing Column Sheet no./rev. 2 Calc. by N Date 3/27/2020 Chk'd by Date App'd by Date Soil weight;Fsoil = hsoil  soil = 120 psf Column no.1 loads Dead load in z;FDz1 = 37.6 kips Live load in z;FLz1 = 30.0 kips Snow load in z;FSz1 = 36.0 kips Wind load moment in x;MWx1 = 20.0 kip_ft Wind load moment in y;MWy1 = 32.0 kip_ft Footing analysis for soil and stability Load combinations per ASCE 7-10 1.0D (0.482) 1.0D + 1.0L (0.794) 1.0D + 1.0S (0.857) 1.0D + 0.75L + 0.75S (0.997) 1.0D + 0.6W (0.875) Combination 7 results: 1.0D + 0.75L + 0.75S Forces on foundation Force in z-axis;Fdz = D  A  (Fswt + Fsoil) + D  FDz1 + L  FLz1 + S  FSz1 = 95.7 kips Moments on foundation Moment in x-axis, about x is 0;Mdx = D  (A  (Fswt + Fsoil)  Lx / 2) + D  (FDz1  x1) + L  (FLz1  x1) + S  (FSz1  x1) = 383.0 kip_ft Moment in y-axis, about y is 0;Mdy = D  (A  (Fswt + Fsoil)  Ly / 2) + D  (FDz1  y1) + L  (FLz1  y1) + S  (FSz1  y1) = 191.5 kip_ft Uplift verification Vertical force;Fdz = 95.74 kips PASS - Foundation is not subject to uplift Bearing resistance Eccentricity of base reaction Eccentricity of base reaction in x-axis;edx = Mdx / Fdz - Lx / 2 = 0 in Eccentricity of base reaction in y-axis;edy = Mdy / Fdz - Ly / 2 = 0 in Pad base pressures q1 = Fdz  (1 - 6  edx / Lx - 6  edy / Ly) / (Lx  Ly) = 2.992 ksf q2 = Fdz  (1 - 6  edx / Lx + 6  edy / Ly) / (Lx  Ly) = 2.992 ksf q3 = Fdz  (1 + 6  edx / Lx - 6  edy / Ly) / (Lx  Ly) = 2.992 ksf q4 = Fdz  (1 + 6  edx / Lx + 6  edy / Ly) / (Lx  Ly) = 2.992 ksf Minimum base pressure;qmin = min(q1,q2,q3,q4) = 2.992 ksf Maximum base pressure;qmax = max(q1,q2,q3,q4) = 2.992 ksf Allowable bearing capacity Allowable bearing capacity;qallow = qallow_Gross = 3 ksf qmax / qallow = 0.997 PASS - Allowable bearing capacity exceeds design base pressure Project Job Ref. Section Footing at Existing Column Sheet no./rev. 3 Calc. by N Date 3/27/2020 Chk'd by Date App'd by Date FOOTING DESIGN (ACI318) In accordance with ACI318-11 incorporating Errata as of August 8, 2014 Material details Compressive strength of concrete;f’c = 4000 psi Yield strength of reinforcement;fy = 60000 psi Cover to reinforcement;cnom = 3 in Concrete type;Normal weight Concrete modification factor; = 1.00 Column type;Concrete Analysis and design of concrete footing Load combinations per ASCE 7-10 1.4D (0.369) 1.2D + 1.6L + 0.5Lr (0.653) 1.2D + 1.6L + 0.5S (0.780) 1.2D + 1.0L + 1.6S (0.931) 1.2D + 1.6Lr + 0.5W (0.302) Combination 6 results: 1.2D + 1.0L + 1.6S Forces on foundation Ultimate force in z-axis;Fuz = D  A  (Fswt + Fsoil) + D  FDz1 + L  FLz1 + S  FSz1 = 143.1 kips Moments on foundation Ultimate moment in x-axis, about x is 0;Mux = D  (A  (Fswt + Fsoil)  Lx / 2) + D  (FDz1  x1) + L  (FLz1  x1) + S  (FSz1  x1) = 572.4 kip_ft Ultimate moment in y-axis, about y is 0;Muy = D  (A  (Fswt + Fsoil)  Ly / 2) + D  (FDz1  y1) + L  (FLz1  y1) + S  (FSz1  y1) = 286.2 kip_ft Eccentricity of base reaction Eccentricity of base reaction in x-axis;eux = Mux / Fuz - Lx / 2 = 0 in Eccentricity of base reaction in y-axis;euy = Muy / Fuz - Ly / 2 = 0 in Pad base pressures qu1 = Fuz  (1 - 6  eux / Lx - 6  euy / Ly) / (Lx  Ly) = 4.471 ksf qu2 = Fuz  (1 - 6  eux / Lx + 6  euy / Ly) / (Lx  Ly) = 4.471 ksf qu3 = Fuz  (1 + 6  eux / Lx - 6  euy / Ly) / (Lx  Ly) = 4.471 ksf qu4 = Fuz  (1 + 6  eux / Lx + 6  euy / Ly) / (Lx  Ly) = 4.471 ksf Minimum ultimate base pressure;qumin = min(qu1,qu2,qu3,qu4) = 4.471 ksf Maximum ultimate base pressure;qumax = max(qu1,qu2,qu3,qu4) = 4.471 ksf Shear diagram, x axis (kips) 66.4 0 0 -66.4 46.9 -46.9 Project Job Ref. Section Footing at Existing Column Sheet no./rev. 4 Calc. by N Date 3/27/2020 Chk'd by Date App'd by Date Moment diagram, x axis (kip_ft) 132.7 0 132.7 0 101.6 Moment design, x direction, positive moment Ultimate bending moment;Mu.x.max = 101.614 kip_ft Tension reinforcement provided;8 No.5 bottom bars (5.9 in c/c) Area of tension reinforcement provided;Asx.bot.prov = 2.48 in2 Minimum area of reinforcement (10.5.4);As.min = 0.0018  Ly  h = 1.037 in2 PASS - Area of reinforcement provided exceeds minimum Maximum spacing of reinforcement (10.5.4);smax = min(3  h, 18 in) = 18 in PASS - Maximum permissible reinforcement spacing exceeds actual spacing Depth to tension reinforcement;d = h - cnom - x.bot / 2 = 8.688 in Depth of compression block;a = Asx.bot.prov  fy / (0.85  f'c  Ly) = 0.912 in Neutral axis factor;1 = 0.85 Depth to neutral axis;c = a / 1 = 1.073 in Strain in tensile reinforcement (10.3.5);t = 0.003  d / c - 0.003 = 0.02130 PASS - Tensile strain exceeds minimum required, 0.004 Nominal moment capacity;Mn = Asx.bot.prov  fy  (d - a / 2) = 102.072 kip_ft Flexural strength reduction factor;f = min(max(0.65 + (t - 0.002)  (250 / 3), 0.65), 0.9) = 0.900 Design moment capacity;Mn = f  Mn = 91.865 kip_ft Mu.x.max / Mn = 1.106 FAIL - Ultimate moment load exceeds design moment capacity One-way shear design, x direction Ultimate shear force;Vu.x = 46.919 kips Depth to reinforcement;dv = h - cnom - x.bot / 2 = 8.688 in Shear strength reduction factor;v = 0.75 Nominal shear capacity (Eq. 11-3);Vn = 2    (f'c  1 psi)  Ly  dv = 52.747 kips Design shear capacity;Vn = v  Vn = 39.56 kips Vu.x / Vn = 1.186 FAIL - Ultimate shear load exceeds design shear capacity Shear diagram, y axis (kips) 66.4 0 0 -66.4 27.5 -27.5 Project Job Ref. Section Footing at Existing Column Sheet no./rev. 5 Calc. by N Date 3/27/2020 Chk'd by Date App'd by Date Moment diagram, y axis (kip_ft) 66.4 0 66.4 0 37.3 Moment design, y direction, positive moment Ultimate bending moment;Mu.y.max = 37.327 kip_ft Tension reinforcement provided;7 No.5 bottom bars (14.8 in c/c) Area of tension reinforcement provided;Asy.bot.prov = 2.17 in2 Minimum area of reinforcement (10.5.4);As.min = 0.0018  Lx  h = 2.074 in2 PASS - Area of reinforcement provided exceeds minimum Maximum spacing of reinforcement (10.5.4);smax = min(3  h, 18 in) = 18 in PASS - Maximum permissible reinforcement spacing exceeds actual spacing Depth to tension reinforcement;d = h - cnom - x.bot - y.bot / 2 = 8.062 in Depth of compression block;a = Asy.bot.prov  fy / (0.85  f'c  Lx) = 0.399 in Neutral axis factor;1 = 0.85 Depth to neutral axis;c = a / 1 = 0.469 in Strain in tensile reinforcement (10.3.5);t = 0.003  d / c - 0.003 = 0.04854 PASS - Tensile strain exceeds minimum required, 0.004 Nominal moment capacity;Mn = Asy.bot.prov  fy  (d - a / 2) = 85.314 kip_ft Flexural strength reduction factor;f = min(max(0.65 + (t - 0.002)  (250 / 3), 0.65), 0.9) = 0.900 Design moment capacity;Mn = f  Mn = 76.783 kip_ft Mu.y.max / Mn = 0.486 PASS - Design moment capacity exceeds ultimate moment load Footing geometry factor (15.4.4.2);f = Lx / Ly = 2.000 Area of reinf. req. for uniform distribution (CRSI);Asreq = (Mu.y.max / (f  fy  (d - a / 2)))  2  f / (f + 1) = 1.407 in2 PASS - Reinforcement can be distributed uniformly One-way shear design, y direction Ultimate shear force;Vu.y = 27.477 kips Depth to reinforcement;dv = h - cnom - x.bot - y.bot / 2 = 8.062 in Shear strength reduction factor;v = 0.75 Nominal shear capacity (Eq. 11-3);Vn = 2    (f'c  1 psi)  Lx  dv = 97.904 kips Design shear capacity;Vn = v  Vn = 73.428 kips Vu.y / Vn = 0.374 PASS - Design shear capacity exceeds ultimate shear load Two-way shear design at column 1 Depth to reinforcement;dv2 = 8.375 in Shear perimeter length (11.11.1.2);lxp = 20.375 in Shear perimeter width (11.11.1.2);lyp = 20.375 in Project Job Ref. Section Footing at Existing Column Sheet no./rev. 6 Calc. by N Date 3/27/2020 Chk'd by Date App'd by Date Shear perimeter (11.11.1.2);bo = 2  (lx1 + dv2) + 2  (ly1 + dv2) = 81.500 in Shear area;Ap = lx,perim  ly,perim = 415.141 in2 Surcharge loaded area;Asur = Ap - lx1  ly1 = 271.141 in2 Ultimate bearing pressure at center of shear area;qup.avg = 4.471 ksf Ultimate shear load;Fup = D  FDz1 + L  FLz1 + S  FSz1 + D  Ap  Fswt + D  Asur  Fsoil - qup.avg  Ap = 120.619 kips Ultimate shear stress from vertical load;vug = max(Fup / (bo  dv2),0 psi) = 176.715 psi Column geometry factor (11.11.2.1); = ly1 / lx1 = 1.00 Column location factor (11.11.2.1);s =40 Concrete shear strength (11.11.2.1);vcpa = (2 + 4 / )    (f'c  1 psi) = 379.473 psi vcpb = (s  dv2 / bo + 2)    (f'c  1 psi) = 386.457 psi vcpc = 4    (f'c  1 psi) = 252.982 psi vcp = min(vcpa,vcpb,vcpc) = 252.982 psi Shear strength reduction factor;v = 0.75 Nominal shear stress capacity (Eq. 11-2);vn = vcp = 252.982 psi Design shear stress capacity (Eq. 11-1);vn = v  vn = 189.737 psi vug / vn = 0.931 PASS - Design shear stress capacity exceeds ultimate shear stress load 1 8 No.5 bottom bars (5.9 in c/c) 7 No.5 bottom bars (14.8 in c/c)